School of Civil and Environmental Engineering
Structural Engineering, Mechanics and Materials Research Report No. 05-1
High Strength/High Performance Concrete for Precast Prestressed Bridge Girders in Georgia
Final Report
Prepared for Office of Materials and Research Georgia Department of Transportation GDOT Research Project No. 9510
by Lawrence F. Kahn, James S. Lai, Aziz Saber, Mohamed Shams,
Chris Reutlinger, Jason Dill, Adam Slapkus, Scott Canfield, and Mauricio Lopez
March 2005
Contract Research GDOT Research Project No. 9510
High Strength/High Performance Concrete for Precast Prestressed Bridge Girders in Georgia
Final Report
Prepared for Office of Materials and Research Georgia Department of Transportation
by Lawrence F. Kahn, James S. Lai, Kimberly Kurtis, Aziz Saber, Mohamed Shams, Chris Reutlinger, Jason Dill, Adam Slapkus, Scott Canfield, Mauricio Lopez and
Bobby Haines
March 2005
The contents of the report reflect the views of the authors who are responsible for the facts and the accuracy of the data presented herein. The contents do not necessarily reflect the official views or policies of the Georgia Department of Transportation. This report does not constitute a standard, specification, or regulation.
Executive Summary
The research project found that by using high strength/high performance concrete (HPC) precast prestressed bridge girders could be fabricated at precast concrete plants and that those girders could be used for construction of economical, long-span bridges. The research was divided into seven tasks. In the first, simple span composite highway bridges were designed using girder strengths varying from 6,000 psi to 15,000 psi, composite deck strengths varying from 3,500 psi to 7,000 psi, girder spacing varying from 5 ft to 11 ft, and using prestressing strand diameters of 0.5-in., 0.5-in. special, and 0.6-in. AASHTO Type I, II, III and IV girders were considered. The maximum effective concrete strength was found to be 13,000 psi. Using this strength concrete in combination with 0.6-in. diameter prestressing strands, the maximum span length of bridges could be increased 40 percent compared to bridges designed with 6,000 psi girders using 0.5-in strand at the same girder spacing.
The materials studies developed HPC mixes using cement, Class F fly ash, silica fume and high range water reducing admixtures. Mixes made in the laboratory, in ready-mix trucks, and at precast plants showed that HPC with design strengths of 7,000 psi, 10,000 psi and 14,000 psi and with low permeabilities (less than 2,000 coulombs) could be made using locally available fine and coarse aggregates. Both limestone and granite-gneiss coarse aggregates from across Georgia were used to make all grades of HPC. Long-term creep and shrinkage studies showed that the creep of 10,000 to 14,000 psi HPC's was between one-third to one-half that of normal strength concrete and that the shrinkage of those HPC's was between one-half to twothirds that of normal strength concrete. Long-term prestress losses in bridge girders were similarly reduced.
The transfer length (lt) of the 0.6-in. strand was measured on AASHTO Type II and Type IV girders. The values ranged from 14.6 in. on the 14,000 psi Type II girders to 19.4 in. on the 10,000 psi Type IV girders. These transfer lengths were more than 35 percent less than that specified in AASHTO bridge design standards.
The development length (ld) of the 0.6-in. strand was determined by flexure and shear tests of 34-ft long Type II HPC girders. The ld was found to be 80 in. which was less than the
ii
AASHTO specified 96 in. The shear tests showed that current bridge design standards in both the AASHTO Standard 17th Edition and in the AASHTO LRFD 3rd Edition conservatively predicted the shear capacity of HPC bridge girders so long as the maximum stress in the reinforcing bar stirrups was limited to 60,000 psi as given in those codes.
One Type IV and one Bulb-T 56-in. deep girders were constructed using 10,150 psi (70 MPa) HPC; each girder was 89 ft (27.1 m) long. The Type IV used 52 0.6-in. (15 mm) prestressing strands while the BT-56 used 44 0.6-in. (15 mm) strands. Flexural tests of those girders and the flexural tests of the Type II girders made with 10,000 psi (69 MPa) and 14,000 psi (96 MPa) design strength HPC showed that the actual moment strength of each girder was greater than the strength predicted using the AASHTO Standard Specifications for Highway Bridges, 16th and 17th Editions. The load-deflection behavior of the girders was accurately predicted using theoretical moment-curvature relations. Cracking moments were accurately predicted using a maximum tensile stress of 7.5fc' pursuant to the AASHTO Standard Specifications.
A high performance concrete, four span highway bridge was designed by the Georgia Department of Transportation as a demonstration HPC bridge in Henry County, Georgia. The design included Type II girders with a 50-ft 1-in. (15.3 m) span and Type IV girders with a 124-ft 1-in. (37.8 m)span. The girders had a design strength of 10,150 psi (70 MPa). The HPC 8-in. thick composite deck had a design strength of 7,250 psi (50 MPa). The construction provided girders with a 56-day mean compressive strength over 13,000 psi (89.6 MPa) and chloride permeability less than 200 coulombs (required permeability was less than 3000 coulombs). The mean deck strength for the longest span was 6,800 psi (46.9 MPa) with a chloride permeability of 4,300 coulombs; these deck quality values showed a performance less than required. The HPC bridge was constructed at an approximate cost of $51.38 per square foot of finished bridge deck, compared with the average cost of $56.65 for a typical, normal strength concrete, precast prestressed composite bridge in Georgia.
Long-term strain and deflection measurements of the demonstration bridge showed that strains at the level of the prestressing strands in the girders due to creep and shrinkage ceased three months after the composite deck was cast. Seasonal changes over the three years of monitoring caused strain variation of less than 40 and deflections less than 0.25 in (6.4 mm). These measurements indicated that the long-term losses in prestressing force were very low.
iii
During a one month period after the composite deck was placed, thermal and shrinkage contraction of the deck caused an increased deflection of 0.58 in. ((14.7 mm) for the 124-ft (37.8 m) span girder; this deflection equaled 22 percent of the deflection due to the weight of the deck alone. Similar deck-contraction induced deflections were noted in the laboratorytested Type IV and BT-56 girders. This phenomenon should be considered when computing girder deflections of long span bridges.
It is recommended that precast prestressed bridge girders be constructed using high performance concrete and using 0.6-in. (15 mm) diameter, 270 ksi (1862 MPa) prestressing strand. The current AASHTO design specification may be used to conservatively design HPC bridge girders with compressive strengths as high as 14,150 psi (70 MPa).
Flexural test of 10,150 psi, 89-ft long Type IV girder
Erection of 10,150 psi, 124-ft long Type IV girder for the HPC demonstration bridge iv
Acknowledgements
The Federal Highway Administration and the Georgia Department of Transportation sponsored the research reported herein through a pooled-fund project, Georgia DOT research project no. 9-9510-0-97-50550, Task Order no. 64. Standard Concrete Products supplied all precast field concrete and all precast prestressed girders. Tidwell Construction built the Henry County HPC demonstration bridge. These contractors were always helpful and of great assistance. For laboratory phases of the overall research project, LaFarge Cement, Boral Material Technologies, and Grace Construction Products donated cement, flyash, and concrete admixtures, respectively. The support provided by the sponsors is gratefully acknowledged.
The findings and conclusions reported herein are those of the authors and do not necessarily represent the opinions, conclusions, or specifications of the Federal Highway Administration, the Georgia Department of Transportation, or any other sponsoring or cooperating organization.
The following individuals from the Georgia Department of Transportation were extremely helpful with the research: Mr. Paul Liles, Ms. Lyn Clements, Mr. Rick Deaver, Ms. Supriya Kamatkar, Mr. Myron Banks, and Ms. Melissa Harper. Ms. Marcia Simon of the Federal Highway Administration gave many valuable suggestions concerning the development of this research. The following Georgia Tech students assisted with experimental phases of the study: Brandon Buchberg, Natalie Hodges, Richard Jennings, Alan Mullenix, K. Fred Meyer, Brooke Ramage, and Maria Wilmhof. Their assistance is gratefully acknowledged.
v
Table of Contents
Page
Executive Summary
ii
Acknowledgements
v
Notation
viii
Chapter
1. Introduction
1-1
1.1
Purpose and Objectives
1-1
1.2
Scope
1-2
1.3
Organization
1-2
2. Analytical Investigation
2-1
2.1
Introduction
2-1
2.2
Maximum Girder Lengths
2-2
2.3
Stability Considerations
2-10
3. HPC Mix Designs and Properties
3-1
3.1
Introduction
3-1
3.2
Materials
3-1
3.3
Grade 1 HPC
3-3
3.4
Grade 2 HPC
3-10
3.5
Grade 4 HPC
3-24
3.6
Creep and Shrinkage Properties of HPC
3-30
4. Type II Girders: Transfer and Development Length of 0.6-in. Strand
4-1
4.1
Introduction
4-1
4.2
Type II Girder Design, Construction, and Instrumentation
4-2
4.3
Material Properties
4-5
4.4
Transfer Length
4-7
4.5
Girder Test Results
4-9
4.5.1 Flexural Behavior
4-11
4.5.2 Shear Behavior
4-12
4.5.3 Bond Behavior and Development Length
4-15
4.6
Comparison with Other Research
4-17
4.7
Conclusions Regarding Transfer and Development Length
4-17
vi
5. HPC Demonstration Bridge Evaluation
5-1
5.1
Introduction
5-1
5.2
Girder Construction and Instrumentation
5-1
5.3
Transfer Length
5-11
5.4
Deck Construction
5-12
5.5
Material Evaluation
5-15
5.6
Load Test on Phase 1
5-24
5.7
Long-term Bridge Performance
5-28
6. Type IV and Bulb-T 56 Flexural Test Results
6-1
6.1
Introduction
6-1
6.2
Girder Design
6-1
6.3
Instrumentation
6-6
6.4
Girder Construction
6-9
6.5
Material Properties
6-12
6.6
Deck Induced Girder Deflections
6-18
6.7
Transfer Length
6-23
6.8
Flexural Test Results and Behavior
6-28
7. Conclusions and Recommendations
7-1
7.1
Conclusions
7-1
7.2
Recommendations for Design Implementation
7-3
7.3
Recommendations for Future Research
7-4
References
R-1
vii
Notation
Report a
ACI 3182005 --
Ac
--
Anc
--
AASHTO --
Aps
Aps
Apse
--
Av
Av
BCL
--
CM
--
CSS
--
DAQ
--
D
db
DEMEC --
dp
dp
Ec
Ec
Eci
--
Eps
fc'
fc'
fci'
fci'
fd
fd
fcds
fr
fr
fpc
fpc
fpe
fpe
fps
fps
AASHTO Standard
Description
--
Shear span, distance from support to point load on girder
--
Cross sectional area of composite girder (combined area of girder and deck)
--
Cross sectional area of girder
--
American Association of State Highway and Transportation Officials
As*
Cross sectional area of prestressing strand
--
Effective area of prestressed reinforcement adjusted inside the transfer or development length regions
Av
Area of shear (stirrup) reinforcement
--
Distance from bottom of girder to center line of bottom row of prestressing strands
--
Cementitious Materials
--
Concrete Surface Strain
--
Data Acquisition
D
Diameter of prestressing strand
--
Detachable Mechanical Strain Gage
d
Distance from compression fiber to centroid of prestressed reinforcement
Ec
Modulus of elasticity of concrete based on 6 x 12 cylinder
--
Modulus of elasticity of concrete based on 6 x 12 cylinder at strand release
Elastic modulus of prestressing steel (ksi)
fc'
Concrete compressive strength at specified time
fci'
Concrete compressive strength at strand release
Tensile stress at the extreme tensile fiber due to the dead load
of the girder and slab
stress in concrete at the cgs due to all superimposed dead
loads (ksi)
fr
Modulus of rupture of concrete
Resultant compressive stress at the centroid of the composite
fpc
section or at the junction of the web and flange when the centroid lies within the flange due to both prestress and
moments resisted by the precast member acting alone.
Compressive stress at the extreme tensile fiber due to effective
prestressing force
fsu*
Stress in prestressed reinforcement at nominal strength of member
viii
Report
fpt fpu fr fse fsi fsu fy h Ic Inc
ld
lfb
lt LOLAX MDL Mcr Mmax Mn Mu n
R2
s S Stop-nc Sbot-nc Stop-c Sbot-c t
to
ACI 3182005 -fpu
fse --
fy h ---
ld
--
----
Mn Mu n
s
-----
AASHTO Standard
Description
--
Stress in prestressing strand just prior to strand release
fs'
Specified tensile strength of prestressed reinforcement
modulus of rupture of concrete
fse
Effective prestressing stress after losses
--
Stress in prestressing strand just after strand release
Stress in prestressed reinforcement at nominal strength of
member
fsy
Specified yield strength of non-prestressed reinforcement
h
Overall depth of the composite girder; Also used as relative humidity for shrinkage calculations
--
Moment of inertia of composite girder (girder and deck)
--
Moment of inertia of girder
Development length of prestressing strand (In AASHTO
--
Standard, ld refers to non-prestressed reinforcement
development length)
Flexural bond length. Additional length over which the strand
--
should be bonded so the stress fps may develop in the strand at
the nominal strength of the member.
--
Transfer length of prestressing strand
--
Low relaxation loss prestressing strand
--
Moment due to Dead Load
Cracking moment
Maximum observed experimental moment
Mn
Nominal moment strength
Mu
Ultimate moment strength = Mn
--
Modular Ratio
Coefficient of determination; it measures the quality of the regression analysis. It is equal to one for a perfect fit and to
zero for unrelated data and model.
s
Spacing of shear reinforcement
Beam or specimen surface area (in2)
--
Top section modulus for non-composite girder
--
Bottom section modulus of non-composite girder
--
Top section modulus for composite girder
--
Bottom section modulus for composite girder
Age of concrete (days) for creep and shrinkage calculations
Age of concrete at loading (days) for creep and shrinkage
calculations
TCL
--
--
Distance from top of girder to center line of top row of prestressing strands
ix
Report
W/CM wc V Vc
Vci
Vcw
Vd Vs Vp VWSG sh cu ps psmax su t u
ACI 3182005 -wc Vc Vci
Vcw Vd Vs Vp --
AASHTO Standard
Description
--
Water to Cementitious Materials Ratio
wc
Unit weight of concrete
Beam or specimen volume (in3)
Vc
Nominal shear strength provided by concrete
Vci
Nominal shear strength provided by concrete when diagonal cracking results from combined shear and moment
Nominal shear strength provided by concrete when diagonal
Vcw
cracking results from excessive principal tensile stress in the
web
Vd
Shear force from dead load of girder plus deck slab
Vs
Nominal strength provided by shear reinforcing steel
Vp
Vertical component of the effective prestressing force
--
Vibrating Wire Strain Gage
Shrinkage constant depending on member shape and size
Shrinkage strain
Ultimate concrete compressive strain
Strain in prestressing strands at nominal strength
Maximum observed strain in prestressing strands
Breaking tensile strain of prestressing strand
Creep coefficient parameters (ACI-209)
Correction factor of unit weight of concrete, taken as 0.85
Stress-to-strength ratio at loading for creep calculations
Creep coefficient at age "t" loaded at t
Ultimate creep coefficient
Creep constant depending on member shape and size
x
Chapter 1. Introduction
1.1 Purpose and Objectives The purpose of the research was to investigate the viability of using high-strength/high-
performance concrete in precast prestressed concrete bridges in Georgia and in the nation. Several specific objectives were included in determining whether high performance concretes (HPC) could be effectively utilized.
The first objective was to determine analytically if HPC could be used with existing prestressed beam types (e.g. AASHTO Type I, II, III and IV) to extend their usable and effective span length and to increase the spacing between the girders. Such extension would permit the lengthening of existing bridges without the need to modify embankments, approaches, and other roadway geometrics.
A second objective was to develop mix proportions for the Grades 1, 2, and 4 HPC with 7,000 psi, 10,000 psi and 14,000 psi 56-day target compressive strengths, respectively, using locally available aggregates and to evaluate the properties of the HPC produced. This objective includes design of HPC with rapid strength gain to permit release of prestressing forces within 24 hours after concrete placement.
The third objective is to determine if concrete plants in Georgia have the capability to produce HPC with required quality and to assess the variability of producing the HPC by the local ready-mix concrete plants and the precast concrete plants.
The fourth objective was to evaluate the transfer and development length of 0.6-in. diameter prestressing strands in the prestressed concrete members. Flexure and shear tests of AASHTO Type II girders made with HPC were used to meet this objective along with direct pull-out tests (Mustafa tests) of the 0.6-in. strand.
The fifth objective was to design, construct and evaluate a demonstration bridge made using HPC Type II and Type IV girders and an HPC composite deck. The short-term behavior and long-term performance of this HPC demonstration bridge would be used to validate the use of HPC.
1-1
The sixth objective was to test both an AASHTO Type 4 and a Bulb-T 56-in. deep girder made of identical HPC as used in the demonstration bridge in order to validate the use of current design standards for predicting the strength of such full-size pretensioned girders.
1.2 Scope The scope of the research was limited in the following manners. First, only aggregates
and cements available in Georgia were used to make the laboratory and field batches of HPC. Second, only four Type II, one Type IV and one Bulb-T girders were fabricated for laboratory testing in order to limit expenditures. Third, a single demonstration bridge was constructed and monitored for a period of three years.
1.3 Report Organization The report is organized into chapters which mimic the objectives of the research.
Chapter 2 presents the analytical investigation and concentrates on the increase of span lengths when the concrete compressive strength (fc') increases from the standard 6,000 psi to the maximum of 15,000 psi. Chapter 3 discusses the results of the laboratory mix design development and of the field tests performed at local precast concrete plants. Long-term results from creep and shrinkage studies include results from laboratory, field and demonstration bridge concretes. Chapter 4 presents the results of the Type II girder tests where transfer length of the 0.6-in. strand and the development length of those strands were determined. Chapter 5 details the design, construction and evaluation of the demonstration bridge. The long-term results include the girder strain and deflection over a three year period. Chapter 6 compares the predicted flexural strength of both an HPC Type IV and a Bulb-T 56-in. girder with their actual strength. Chapter 7 gives conclusions and recommendations as to the applicability of using HPC for precast prestressed bridge girders and for cast-in-place bridge decks.
1-2
Chapter 2. Analytical Investigation
2.1 Introduction
The design procedure used in this study conforms to the current AASHTO 16th Edition Specification (1996), using both allowable stress and ultimate strength design criteria. The specified highway bridge loading was HS 20-44 and the military loading which were used along with the Georgia Department of Transportation computer program PCPSBM1R to perform the design calculations. The computer program was modified to permit the use of 0.6in. strands, high performance concrete with strengths up to 20,000 psi, and allowable stress changes from the 15th to 16th Editions of the AASHTO Standard Specifications (1992 and 1996).
The study was focused on pretensioned concrete girders that were part of non-skew simple span bridges. The bridge deck concrete strength was 3.5 ksi, and the thickness was 7 in. The composite deck was considered to be applied with the girders unshored.
AASHTO Type I, II, III, and IV girder sections were used with girder spacing (GS) of 5, 7, 9 and 11 ft. Girder concrete design strengths (56-day) varied from 6 to 15 ksi. The concrete strength at release was assumed as 75 percent of the design strength.
All designs used low relaxation seven wire strands of 0.5-in. and 0.6-in. diameters placed in 2-in. grid. The 0.5-in. strands were the modified strands with cross sectional area equal to 0.167 square in.. The 0.6-in. strands have cross sectional area equal to 0.217 square in.. The ultimate capacity for the two types of strands was 270 ksi. Allowable stresses in prestressing steel were according to AASHTO article 9.15.1. Draping was used to control stresses at the end of members. In two special studies presented in Chapter IV, 0.5-in. strand 270 ksi and 0.5-in. 300 ksi were investigated.
The HS 20-44 live load (Truck, Lane and Military), according to AASHTO article 3.7, and Impact Load according to AASHTO article 3.8, were used. A future wear surface 2-in. thick was taken as an additional dead load. The load distribution factors for moment and shear were considered according to AASHTO article 3.22.
Concrete allowable stresses were according to AASHTO article 9.15
2-1
1. Initial Tension 7.5fci' 2. Final Tension 6fc' 3. Initial Compression 0.6 fci' 4. Final Compression 0.6 fc' 5. Compressive Stresses due to effective prestress plus permanent (dead) loads were
limited to 0.4 fc' 6. Compressive Stresses due to live load plus one half the compressive stresses due to
effective prestress plus permanent (dead) loads were limited to 0.4 fc'.
2.2 Maximum Girder Lengths
The results of the parametric study of AASHTO Type I through IV sections are presented in Figures 2.1 through 2.5. The design of prestressed girders using 0.6-in. strands generally was controlled by the compressive stress conditions at strand release. The initial compression stress condition also was controlling the design case for wider girder spacing using 0.5-in. modified prestressing strands. The large prestressing force that was required to be placed in the girder during construction so as to resist the service loads, caused the allowable stresses at strand release to control the design.
At final conditions the allowable tensile stresses controlled the design of longer spans with close girder spacing using 0.5-in. strands. In cases for very long maximum spans and 5-ft. girder spacing, the compression stress in top of the beam due to effective prestress plus permanent (dead ) load controlled the design.
The results of this study indicated that for a given span length, an increase in concrete strength would increase the allowable stresses which resulted in a decrease in the required number of prestressing strands. An increase in concrete strength allowed the use of larger prestressing force that resulted in longer spans. The larger prestressing force could be achieved by using larger size strands, increasing the total number of strands, or by increasing the ultimate strength of the strands. The increase in the total number of strands would reduce the eccentricity of the prestressing force; thus increasing the number of strands tended to be the least effective option.
2-2
The results of the many designs showed an increase in maximum span with increasing concrete strength for the sections and girder spacing considered. The increase in span when concrete strength was increased from 6 ksi to 14 ksi ranged from 20 to 45 percent. The increase in span when concrete strength was increased from 6 ksi to 9 ksi ranged from 15 to 25 percent. As an example, Table 2.1 lists the span length increase for each girder type for 7-ft girder spacing.
Table 2.1 Span Length Increase for 7-ft Girder Spacing.
Girder Section
Type I Type II Type III Type IV NU1100 NU1350
f'c from 6 ksi to 9 ksi
0.6-in.
0.5-in. Modified
Strands
Strands
24%
18%
20%
20%
19%
19%
19%
19%
20%
20%
24%
22%
f'c from 6 ksi to 14 ksi
0.6-in.
0.5-in. Modified
Strands
Strands
45%
27%
40%
32%
33%
23%
36%
26%
36%
23%
35%
25%
As the strength of concrete was increased, its further effectiveness in increasing the span length diminishes. As shown in Figure 2.4 for Type IV girders using 0.6-in. strand, the gain in span length for fc' greater than 12 ksi was less than 5 percent. The effect was more pronounced for girders at close spacing.
The use of 0.6-in. strand permitted the center of gravity for the strands to be lower for the same total area of steel compared to using 0.5-in. strand. Within the zone below the girder neutral axis, the same total number of strands could be placed, but the 0.6-in. strand yields 30% greater area than the 0.5-in. modified strand.
High strength concretes were more fully utilized with 0.6-in. strand because the use of 0.6-in. strands provided higher prestressing force than 0.5-in. modified strands. The compression stress at strand release generally controlled the design for maximum span lengths using 0.6-in. strands. While the final tension controlled most of the design cases for maximum span lengths using 0.5-in. modified strands. Figure 2.5 illustrates a typical comparison of 0.6in. and 0.5-in. modified strand.
2-3
Max. Span (m)
Max. Span (ft)
Max. Spans AASHTO TYPE I
(Strands 0.5" modified - 270ksi) Girder Concrete Strength (MPa)
41 48 55 62 69 76 83 90 96 103 110
200
60
180
Girder Spacing 11ft
Girder Spacing 9ft 54
160
Girder Spacing 7ft
Girder Spacing 5ft 48
140
42
120
36
100
30
80
24
60
18
40
12
20
6
6 7 8 9 10 11 12 13 14 15 16
Girder Concrete Strength (ksi)
Max. Spans AASHTO TYPE I (Strands 0.6"- 270ksi)
Girder Concrete Strength (MPa)
41 48 55 62 69 76 83 90 96 103 110
200
60
180
Girder Spacing 11ft
Girder Spacing 9ft 54
160
Girder Spacing 7ft
Girder Spacing 5ft 48
140
42
120
36
100
30
80
24
60
18
40
12
20
6
6 7 8 9 10 11 12 13 14 15 16
Girder Concrete Strength (ksi)
Max. Span (m)
Max. Span (ft)
Figure 2.1 Maximum Span Length for AASHTO Type I Girders. 2-4
Max. Span (m)
Max. Span (ft)
Max. Spans AASHTO TYPE II (Strands 0.5" modified - 270ksi)
Girder Concrete Strength (MPa)
41 48 55 62 69 76 83 90 96 103 110
200
60
180
Girder Spacing 11ft
Girder Spacing 9ft 54
160
48
140
Girder Spacing 7ft
Girder Spacing 5ft 42
120
36
100
30
80
24
60
18
40
12
20
6
6 7 8 9 10 11 12 13 14 15 16
Girder Concrete Strength (ksi)
Max. Spans AASHTO TYPE II (Strands 0.6"- 270ksi)
Girder Concrete Strength (MPa)
41 48 55 62 69 76 83 90 96 103 110
200
60
180
Girder Spacing 11ft
Girder Spacing 9ft 54
160
Girder Spacing 7ft
Girder Spacing 5ft 48
140
42
120
36
100
30
80
24
60
18
40
12
20
6
6 7 8 9 10 11 12 13 14 15 16
Girder Concrete Strength (ksi)
Max. Span (m)
Max. Span (ft)
Figure 2.2 Maximum Span Length for AASHTO Type II Girders. 2-5
Max. Span (m)
Max. Span (ft)
Max. Spans AASHTO TYPE III (Strands 0.5" modified - 270ksi) Girder Concrete Strength (MPa)
41 48 55 62 69 76 83 90 96 103 110
200
60
180
Girder Spacing 11ft
Girder Spacing 9ft 54
160
Girder Spacing 7ft
Girder Spacing 5ft 48
140
42
120
36
100
30
80
24
60
18
40
12
20
6
6 7 8 9 10 11 12 13 14 15 16
Girder Concrete Strength (ksi)
Max. Spans AASHTO TYPE III (Strands 0.6"- 270ksi)
Girder Concrete Strength (MPa)
41 48 55 62 69 76 83 90 96 103 110
200
60
180
Girder Spacing 11ft
Girder Spacing 9ft 54
160
Girder Spacing 7ft
Girder Spacing 5ft 48
140
42
120
36
100
30
80
24
60
18
40
12
20
6
6 7 8 9 10 11 12 13 14 15 16
Girder Concrete Strength (ksi)
Max. Span (m)
Max. Span (ft)
Figure 2.3 Maximum Span Length for AASHTO Type III Girders. 2-6
Max. Span (m)
Max. Span (ft)
Max. Spans AASHTO TYPE IV (Strands 0.5" modified - 270ksi) Girder Concrete Strength (MPa)
41 48 55 62 69 76 83 90 96 103 110
200
60
180
54
160
48
140
42
120
36
100
30
80 60
Girder Spacing 11ft
Girder Spacing 9ft
24 18
40
Girder Spacing 7ft
Girder Spacing 5ft 12
20
6
6 7 8 9 10 11 12 13 14 15 16
Girder Concrete Strength (ksi)
Max. Spans AASHTO TYPE IV (Strands 0.6"- 270ksi)
Girder Concrete Strength (MPa)
41 48 55 62 69 76 83 90 96 103 110
200
60
180
54
160
48
140
42
120
36
100
30
80
24
60
Girder Spacing 11ft
Girder Spacing 9ft 18
40
Girder Spacing 7ft
Girder Spacing 5ft 12
20
6
6 7 8 9 10 11 12 13 14 15 16
Girder Concrete Strength (ksi)
Figure 2.4 Maximum Span Length for AASHTO Type IV Girders.
Max. Span (m)
Max. Span (ft)
2-7
Max. Span (ft) Max. Span (m)
Max. Spans AASHTO Girders (7-ft Spacing)
Type IV - 0.6" Strds Type III - 0.6" Strds Type II - 0.6" Strds Type I - 0.6" Strds
Type IV - 0.5" Mod. Strds Type III - 0.5" Mod. Strds Type II - 0.5" Mod. Strds Type I - 0.5" Mod. Strds
Girder Concrete Strength (MPa)
41 48 55 62 69 75 82 89 96 103 110
160
72
140
63
120
54
100
45
80
36
60
27
40
18
20
9
0
0
6 7 8 9 10 11 12 13 14 15 16
Girder Concrete Strength (ksi)
Figure 2.5 Effect of Strand Size on Maximum Span Length for AASHTO Girders (GS=7ft).
2-8
Deflections for girders at maximum spans were compared to AASHTO article 9.11.3 provisions (live load deflection less than L/800). In most of the cases presented here, the requirements of AASHTO were met for all girder types, for each strand type, and for most girder spacings. The few design cases where the girder deflection exceeded the AASHTO allowable were limited to the longest span girders using 0.6-in. strand where the girders were spaced at 5 ft.
The maximum effective girder compressive strengths presented in Table 2.2 were based on the results for the maximum spans for the designs discussed in this chapter.
The maximum effective strength was determined based on the following criteria: 1. The concrete strength at which the slope of the maximum span curve ceased to
increase. 2. The concrete strength where the maximum span length was at least 95% of the
maximum length reached using 15-ksi concrete. 3. The concrete strength where a 1,000 psi increase in concrete strength produced a
change in span length of less than 2%. Based on the above conditions, the maximum effective concrete strength listed in Table 2.2 was selected at the value where the increase in the concrete strength ceased to effectively increase the maximum achievable spans. The findings for 0.5-in. modified strands generally agree with the conclusion by Russell et al. (1995) that the maximum effective concrete strength is 10,000 psi.
2-9
Table 2.2 Maximum Effective Girder Compressive Strength.
Girder Section AASHTO Type I AASHTO Type II AASHTO Type III AASHTO Type IV
Maximum Effective Girder Compressive Strength (ksi)
0.6-in. Diameter Strands
0.5-in. Diameter Modified Strands
11 ft 9 ft
7 ft
5 ft 11 ft 9 ft
7 ft
5 ft
Spacing Spacing Spacing Spacing Spacing Spacing Spacing Spacing
13
12
12
12
11
11
10
10
13
13
13
12
11
11
11
11
12
12
12
11
11
10
10
9
12
12
12
12
11
10
10
10
2.3 Stability Considerations
The stability of long prestressed girders was investigated first for girder stability during bridge deck construction and second for girder stability during handling and erection.
One of the critical loading cases for prestressed girders is when the cast-in-place deck is constructed and before it gained its design strength. During this construction stage, the girder is subjected to axial and transverse loads applied in the plane of its web, where the flexural rigidity is many times greater than the lateral rigidity. If the prestressed girder is supported at the ends and if lateral supports are only located at the ends, this girder could buckle and collapse before it could have achieved its ultimate strength. The axial load is the prestressing force; the transverse load is the girder self weight, the deck weight and any gravity load on the deck.
The lateral stability of long prestressed girders was investigated based on the theory of lateral torsional buckling. The discussions on members subjected to buckling due to lateral bending, twisting and warping by N. S. Trahair (1993), provided some of the background for this study.
The axial critical load for the stability calculations is the total prestressing force. The value of critical load would change as the value of the eccentricity relative to the centroidal axis of the section changed. The eccentricity of the prestressing force was determined by the
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number of strands used in the girder, and where the strands were placed in the section. The maximum and minimum eccentricities were determined for each girder type. The two extreme eccentricities, along with three different concrete compressive strengths ( 6 ksi, 9 ksi and 14 ksi) were used for the critical axial load in Figure 2.6 for an AASHTO Type IV girder. As the girder gets longer, a lower prestressing force would induce lateral stability failure. If the critical axial force approached the actual prestressing force needed for a girder, then the girder potentially would undergo a stability failure during construction.
The results indicated that the critical load and the maximum unsupported length for AASHTO girder sections were increased as the concrete compressive strength increased. The rate of increase in the critical load was almost equal to the rate of increase in the concrete compressive strength. The rate of increase in the unsupported length was much slower than the rate of increase in the concrete compressive strength.
Max. Effective Prestressing Force (Kips)
Prestressing Force as Controlled by Girder Lateral Stability
F'c 6ksi Ecc -22.73" F'c 9ksi Ecc -14.06"
F'c 6ksi Ecc -14.06" F'c 14ksi Ecc -22.73"
F'c 9ksi Ecc -22.73" F'c 14ksi Ecc -14.06"
2000 1800 1600
AASHTO Type IV Deck Wt 91.67Lb/in
1400
1200
1000
800
600
400
200
0 60 70 80 90 100 110 120 130 140 150 160 170 180 190 200 Girder Unsupported Span (Ft)
Figure 2.6 Prestressing Force for Girder Lateral Stability for Type IV Girders.
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The maximum span girders obtained above were investigated for stability during bridge construction. The results were summarized in Table 2.3. In all cases the value of the critical axial load (prestressing force) was higher than the applied force. Therefore the maximum span girders would not buckle during the construction of the bridge deck. The braces were located at distances that exceeded 40 ft, which has been the maximum spacing on lateral brace spacing as recommended by AASHTO 16th Ed. (1996). The brace locations indicated are the maximum spaces required to assure stability.
Table 2.3 Stability of maximum span length girders during bridge deck construction.
AASHTO Type
Girder Spacing 11ft Max. Span Brace Location Applied Force Critical Force
Girder Spacing 9ft Max. Span Brace Location Applied Force Critical Force
Girder Spacing 7ft Max. Span Brace Location Applied Force Critical Force
Girder Spacing 5ft Max. Span Brace Location Applied Force Critical Force
0.6-in. Strands fc' 14 ksi
IV III II I
0.6-in. Strands fc' 9 ksi
IV III II I
123 L/2 2718 2852
94 L/2 2104 2470
66 L/2 1167 2139
48 L/2 865 2572
107 L/2 1659 2942
80 L/2 1106 2611
55 L/2 716 2391
41 L/2 559 2758
135 L/3 2955 5227
103 L/2 2033 2062
73 L/2 1170 1766
54 L/2 1020 2074
116 L/2 1667 2525
89 L/2 1185 2145
63 L/2 795 1873
45 L/2 558 2311
148 L/3 2897 4384
114 L/3 1888 3683
84 L/2 1325 1355
61 L/2 945 1629
130 L/2 1752 2030
102 L/2 1262 1654
72 L/2 793 1445
52 L/2 557 1750
164 L/3 2845 3600
130 L/3 1978 2881
97 L/3 1332 2247
71 L/2 948 1211
145 L/3 1770 3577
120 L/3 1419 2637
85 L/2 793 1047
62 L/2 557 1245
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Chapter 3. HPC Mix Designs and Properties
3.1 Introduction High performance concrete is a relative term. Various agencies, The American
Concrete Institute, the Strategic Highway Research Program and others, have used different definitions to describe high strength and high performance concrete. The definition of HPC could base on mix design ingredients, fresh concrete properties, and long term performance properties. The high performance concrete grades defined by FHWA for the long term performance (Goodspeed et al. 1996) are shown in Table 3-1 along with the standard test methods for the HPC property analysis. Using these grades and the field conditions, a concrete mix designer can specify mix performance criteria to produce a concrete mix that could achieve the expected service life of the concrete structure under the given service conditions.
Table 3.1 Characteristics for performance of high performance concrete grades (Goodspeed et al 1996)
Performance Standard Test
Characteristic
Method
1
FHWA Grade
2
3
4
Strength
AASHTO T 2 41 < x < 55 Mpa 55 < x < 69 Mpa (8000 69 < x < 97 Mpa 97 MPa < x
(x = compressive strength)
ASTM C39
(6000 < x < 8000 psi)
< x < 10000 psi)
(10000 < x < 14000 psi) (14000 < x)
Elasticity (x = modulus of
elasticity)
ASTM 469
28 < x < 40 Gpa 40 < x < 50 Gpa (6000 (4000 < x < 6000 ksi) < x < 7500 ksi)
Chloride AASHTO T 277
Penetraion (x = coulombs)
ASTM C 1202
3000 > x > 2000
2000 > x > 800
Creep
75 > x > 60 / Mpa 60 > x > 45 / Mpa
(x = microstrain/ ASTM C 512 (0.52 > x > 0.41 / psi) (0.41 > x > 0.31 / psi)
pressure unit
Shrinkage (x = microstrain)
ASTM C 157
800 > x > 600
600 > x > 400
50 < x Gpa (7500 ksi < x)
800 > x
45 > x > 30 / Mpa (0.31 > x > 0.21 / psi)
400 > x
30 MPa > x (0.21 / psi > x)
3.2 Materials Type I cement was used for the cast-in-place Grade 1 HPC mixes. Both Type I cement
and Type III cement were used for the Grade 2 and Grade 4 HPC mixes. These two types of cement were used in all the laboratory mixes and in the mixes produced by Blue Circle Ready Mix Concrete, Thomas Ready Mix Concrete and Standard Concrete Plant. Type III cement
3-1
was used in the HPC mixes produced by Allied Ready Mix Concrete and Tindall Concrete Plant.
Table 3.2 summarized the coarse aggregates used for this study. The #67 aggregate from the Blue Circle at Lithonia quarry was used in the laboratory for developing the Grade 1 and Grade 2 HPC mixes and was also used in the later stage in the laboratory for developing the Grade 4 HPC mixes. The #7 aggregate from Blue Circle was used in the laboratory for developing the Grade 4 HPC mixes. The #7 limestone from Florida Rocks was also used in the laboratory at the beginning for developing the Grade 4 HPC mixes.
The fine aggregate used in most of the laboratory HPC mixes as well as used by the concrete plants was a natural sand from the Brown Brothers Junction City quarry. It had a bulk specific gravity of 2.65, and absorption of 0.4% and 0.60% material finer than #200 sieve. Its finest modulus was 2.3.
Table 3.2 Sources of Coarse Aggregate
Quarry Producer Location
General Composition
Agg. Bulk Spec.Gravity Percent
Size Dry
SSD Absorption
Blue
Lithonia Granite/Gneiss #67
2.62 2.63
0.58
Circle
GA
#7
Vulcan Lithonia Granite/Gneiss #67
2.60 2.62
0.58
Materials GA
Southern Postell Mylonite,Gneiss #67
2.62 2.63
0.57
Aggregate GA
Amphibolite
Martin Dan
Mylonite,Gneiss #67
2.64 2.66
0.61
Marietta GA
Amphibolite
Florida Rocks
Rome GA
Limestone
#7
2.68 2.69
0.51
Mineral admixtures used for this project included fly ash and silica fume. Type F fly ash from Monex Resources, conforming to ASTM C 618 was selected. W.R. Grace Force 10,000 densified microsilica fume, conforming to ASTM C 1240 was selected. Type C fly ash was not used in this study because it was not readily available in Georgia.
3-2
Chemical admixtures used for this project included two high range water reducers (HRWR), a set retarder, and an air-entraining admixture (AEA). The high range water reducers included Daracem ML 330, a melamine-based HRWR from W.R. Grace and WRDA-19, a sulfonated naphalene formaldehyde condensated-based HRWR also from W.R. Grace. Both HRWRs met ASTM C 494. The water reducer/set retarder used was WRDA-64 from W.R. Grace, meeting ASTM C 494. The air-entraining admixture used was Darex AEA from W.R. Grace, conforming to ASTM C 260.
3.3 Grade 1 HPC The performance requirements for the Grade 1 HPC mix were:
8,400 psi mean compressive strength at 56-day age (fc' = 7,000 psi). Good durability with a chloride ion penetration between 800 and 2,000 coulombs at 56-
day age 5-8% air content in the fresh concrete Good workability and finishability with the slump between 4 and 6 inches.
To meet the 7,000 psi design strength requirement, an average mix strength of 8,400 psi at 56 day-age was selected. A compressive strength of 1,400 psi over the design strength is specified by the ACI 318-95 Building Code Requirements for Structural Concrete. This criterion was used to establish the compressive strength required for all three grades of HPC concrete mixes in this research study since the standard deviations of the compressive strength of these mixes were not yet available.
The Grade 1 HPC mixes developed in the laboratory are presented in Table 3.3 and 3.4. A total of 12 mixes were developed. Natural sands were used in all Grade 1 mixes to achieve better finishability for the bridge deck construction. The first 2 mix designs used a 50/50 blend of the natural sands from Brown Brothers (BB) and Martin Marietta (MM); all subsequent mixes used 100% Brown Brothers sand. The 12 mixes could be divided into 4 groups.
3-3
Group (1): 734 to 825 lb/yd3 cement, 72 to 100 lb/yd3 fly ash, no silica fume 244 to 307 lb/yd3 water, G1-1 to G1-4
Group (2): 650 to 699 lb/yd3 cement, 97 to 124 lb/yd3 fly ash, 25 lb/yd3 silica fume 232 to 239 lb/yd3 water, G1-5 to G1-7
Group (3): 582 to 642 lb/yd3 cement, 90 to 125 lb/yd3 fly ash, 15 lb/yd3 silica fume 203 to 229 lb/yd3 water, G1-8 to G1-10
Group (4): 594 to 653 lb/yd3 cement, 100 to 148 lb/yd3 fly ash, 15 lb/yd3 silica fume 245 to 251 lb/yd3 water, G1-11 to G1-12
The G1-13 through G1-16 mixes, shown in Table 3.5, were selected for the field production of the Grade 1 HPC mix by a ready-mix concrete truck mixer. These mixes were based on the G1-11 mix. The G1-16 mix was selected as the "best" G1 mix.
The compressive strength testing of the G1-16 mix was performed on the 4-in. diameter cylinder specimens and the 6-in. diameter cylinder specimens at 3-day, 7-day, 28-day, and 56day ages. Results of the compressive strength are summarized in Figure 3.1. The compressive strength of the 4-in. cylinders was between 1 % and 2% greater than that of the 6-in. cylinders. For the G1-16 mix, the 56-day modulus of elasticity (E) using the 6x12 cylinders was 4.21 ksi, the Poison's ratio was 0.19; the modulus of rupture using 4x4x16 in. beams was 840 psi; the rapid chloride ion penetration test (ASTM C-1202) at 56 days gave 1459 coulombs.
To ensure that HPC mixes can be produced statewide using the locally available aggregates, two additional coarse aggregates from two different regions in Georgia were selected for this study. After consulting with the Office of Materials and Research of the GDOT, two #67 aggregates from the following sources were chosen for this study: Martin Marietta's Dan Quarry (outside of Augusta), and Southern Aggregate's Postell Quarry (outside of Macon). These two aggregates together with the #67 from the Blue Circle Lithonia Plant near Atlanta represent the coarse aggregates available in the eastern, central, and northern sections of Georgia.
3-4
Table 3.3 Grade 1 HPC Laboratory mixes (Group 1 and 2)
HPC Project Number: CONSTITUENT INGREDIENTS
Cementitious (lb/yd3) Type I Flyash, Class F Silica Fume
Fine Aggregates (lb/yd3) Natural Sand (MM) Natural Sand (BB)
Coarse Agg. #67 Granite Coarse/Fine Aggregate Ratio Water (Net) (lbs/yd3)
Water / Cement Ratio Water / Cementitious Ratio Chemical Admixture (oz/cwt) : Set Retarder Superplasticizer (NA) Air Entraining (Darex) Total Admixture (lb/cu.yd.) Plastic Concrete Properties Total Wt. /cu yd (lbs) Unit weight (pcf) Slump (in.) Air Contents (%) Compressive Strength (psi) 24 Hour 3 Day 7 Day 28 Day 56 Day Chloride Permeability 28 Days (coulombs) 56 Days (coulombs)
G1-1 G1-2 G1-3 G1-4 G1 - 5 G1 - 6 G1 - 7
823
763
766
734
699
685
650
97
96
72
98
99
90
125
0
0
0
0
25
25
25
394 394 1661 2.11
307 0.37 0.33
462 462 1634 1.77
284 0.37 0.33
0 1002 1636 1.63
239 0.31 0.29
0 950 1759 1.85
244 0.33 0.29
0 1095 1684 1.54
239 0.34 0.29
0 1154 1699 1.47
232 0.34 0.29
0 1146 1700 1.48
232 0.36 0.29
4.49
4.49
4.50
4.00
4.00
4.00
4.00
2.99
4.98 11.20 13.99 34.99 29.99 35.00
1.25
1.61
3.60
3.75
3.45
4.00
3.75
4.94
5.83 10.16 10.96 20.38 17.88 19.09
3680 136.30 5.75 8.80
3707 137.31 3.75 7.75
3725 137.97 6.50 5.25
3796 140.58 6.50 6.50
3861 143.00 5.00 4.75
3902 144.54 6.00 5.50
3897 144.34 5.50 5.25
3040 4305 5025 5940 6545
3460 4440 5040 6070 6545
4915 6495 7680 8465 9295
4560 6155 7335 9250 10,025
6025 8560 9475 12,215 12,785
7210 8590 9630 11,670 12,380
6605 8305 9520 11,770 12,650
N/A
N/A
2138 2493 1119 1128 1173
N/A
N/A
1457
N/A
796
802
834
3-5
Table 3.4 Grade 1 HPC Laboratory mixes (Group 3 and 4)
HPC Project Number: CONSTITUENT INGREDIENTS
Cementitious (lb/yd3) Type I Flyash, Class F Silica Fume
Fine Aggregates (lb/yd3) Natural Sand (MM) Natural Sand (BB)
Coarse Agg. #67 Granite (lb/yd3) Coarse/Fine Aggregate Ratio Water (Net) (lbs/yd3)
Water / Cement Ratio Water / Cementitious Ratio Chemical Admixture (oz/cwt) : Set Retarder Superplasticizer (NA) Air Entraining (Darex)
Total Admixture (lb/yd3) Plastic Concrete Properties Total Wt. /yd3 (lbs) Unit weight (pcf) Slump (in.) Air Contents (%) Compressive Strength psi) 24 Hour 3 Day 7 Day 28 Day 56 Day Chloride Permeability 28 Days (coulombs) 56 Days (coulombs)
G1-8
642 124 15
0 1136 1680 1.48 229 0.36 0.29
3.95 33.10 4.52 18.35
3845 142.39 6.00 4.50
4800 N/A 8017 10092 11034
2173 1231
G1-9
587 98 15
0 1125 1722 1.53 203 0.35 0.29
3.91 35.06 7.29 18.67
3769 139.58 4.00 8.50
4700 N/A 8307 10037 10094
1725 901
G1-10
582 97 15
0 1115 1706 1.53 222 0.38 0.32
3.88 24.23 4.34 12.97
3749 138.83 4.50 5.25
4050 N/A 7146 8922 9678
1986 1274
G1-11
653 100 15
0 1156 1708 1.48 251 0.38 0.33
3.00 15.00 3.75 12.10
3637 144.30 4.50 4.70
4440 5970 6520 8324 8942
N/A 1883
G1 - 12
594 148 15
0 1139 1684 1.48 247 0.42 0.33
4.00 15.00 4.25 12.10
3586 142.24 4.00 5.90
3910 5250 6210 7801 8403
N/A 1736
3-6
Table 3.5 Grade 1 HPC Ready-mix truck mixes. Mixes G1-13 from Allied, G1-14 from Thomas and Mixes G1-15 and -16 from Blue Circle.
HPC Project Number: CONSTITUENT INGREDIENTS Cementitious (lb/yd3) Type I Flyash, Class F Silica Fume Fine Aggregates: Natural Sand (lb/yd3) Coarse Agg. #67 Granite (lb/yd3) Coarse/Fine Aggregate Ratio Water (Net) (lbs/yd3) Water / Cement Ratio Water / Cementitious Ratio Chemical Admixture (oz/cwt) : Set Retarder Superplasticizer (NA) Air Entraining (Darex)
Total Admixture (lb/yd3) Plastic Concrete Properties
Total Wt. /cu yd (lbs) Unit weight (pcf) Slump (in.) Air Contents (%) Compressive Strength psi) 3 Day 7 Day 28 Day 56 Day Chloride Permeability 28 Days (coulombs) 56 Days (coulombs)
G1-13
627 96 15
1110 1641 1.48 231 0.37 0.31
21.7 111 6.00 10.09
3729 138.1 6.5 N/A
5386 6165 6667 6927
N/A N/A
G1-14
604 93 15
1176 1668 1.42 220 0.36 0.31
1870 84.3 4.90 7.71
3783 140.1
4 3.50
3245 4214 5341 5762
N/A N/A
G1-15
603 88 15
1258 1636 1.30 229 0.38 0.33
18 131 11.00 11.42
3840 142.24
4.5 5.25
3968 4926 6512 7348
1986 1274
G1-16
647 100 15
1146 1693 1.48 229 0.35 0.30
22.4 75.3 6.00 7.47
3847 142.1
4 5.50
6328 7093 8868 9641
N/A 1459
3-7
10000
Compressive Strength, ps
9000
8000
7000 4x8 cyl
6000
6x12 cyl
5000 0
7 14 21 28 35 42 49 56 Days
Figure 3.1 Compressive strength of selected G1 mix (G1-16)
Three mixes using the identical mix design were conducted using the coarse aggregates from the Augusta, Macon, and Blue Circle quarries and these three mixes were designated as G1-17, G1-18, and G1-19 respectively (Table 3.6). Their compressive strengths are shown in Figure 3.2.
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Table 3.6 G1 batches from statewide sources
HPC Project Number:
G1-16 G1-17 G1-18
Aggregate Source
Batch Size CONSTITUENT INGREDIENTS Cementitious:
Type I Flyash, Class F Silica Fume Fine Aggregate: Natural Sand Coarse Agg. #67 Granite Water: (Net) (lbs/yd3) Water / Cement Ratio Water / Cementitious Ratio Chemical Admixture (oz/yd) : Set Retarder Superplasticizer (Naphthalene) Air Entraining (Darex) PLASTIC CONCRETE PROPERTIES Unit weight (pcf) Slump (in.) Air Contents (%) COMPRESSIVE STRENGTH (psi) 24 Hour 3 Day 7 Day 28 Day 56 Day
1.0 yd3
650 100 15 1130 1700 230 0.35 0.30
21 75 6
~ 3-3.5 ~ 5-6
6328 7093 8868 9641
Augusta 1.1 ft3
26.5 4.1 0.6 46 69.3 9.4 0.35 0.30
0.9 7.3 0.34
149.7 4 2.2
6911 8761 9338 11271 12631
Macon 1.1 ft3
26.5 4.1 0.6 46 69.3 9.4 0.35 0.30
0.9 6.5 0.74
151.5 2 3.8
7547 8827 9776 11899 13009
G1-19 Blue Circle 1.1 ft3
26.5 4.1 0.6 46 69.3 9.4 0.35 0.30
0.9 6.8 0.67
147.6 2 3.5
7235 8620 9374 11704 13036
It was concluded that Grade 1 HPC mixes intended for use in the cast-in-place bridge decking can be produced utilizing the conventional constituent materials, conventional production and curing procedures to meet the strength and durability requirements. The workability of the Grade 1 HPC mixes produced in this study was judged to be very good, comparable to a typical normal strength air-entrained concrete mix for the bridge deck construction. The field production of the Grade 1 HPC mixes can be successfully done using the conventional ready-mix concrete truck mixer. The compressive strength requirements can be easily achieved for the Grade 1 mixes produced in the laboratory and by the ready-mix concrete truck mixer. Use of silica fume, even only 15 lb/yd3, in the concrete mix was found to significantly improve the impermeability of the HPC mix and to meet this requirement. Three identical Grade 1 HPC mixes were produced using the aggregates available in the eastern,
3-9
central and northern regions of Georgia. The compressive strengths of these three mixes were within 10 % difference at the same testing age. This consistency of the strength indicated that Grade 1 HPC could be produced using these three different aggregates.
Compressive Strength (psi)
14000
13000 12000
11000
10000 9000
8000 7000
6000 0
7 14
G1-17 Augusta
21 28 35 Time (days)
G1-18 Macon
42 49 56 G1-19 Blue Circle
Figure 3.2 Compressive strength for G1-17, G1-18, and G1-19 mixes
3.4 Grade 2 HPC The goal of Grade 2 HPC was to develop concrete with a minimum specified design
strength of 10,000 psi and an average mix strength of 11,400 psi at 56 day-age for precast prestressed concrete bridge girders. Also required for this mix was chloride permeability between 2,000 to 3,000 coulombs, an air content of 5-8 %, and a mix with good workability a slump between 4 and 6 inches.
The materials used for developing the Grade 2 HPC mixes in the laboratory are summarized below:
Cement: Fly ash: Silica Fume:
Type I and Type III cement from Blue Circle Atlanta Plant Type F fly ash from Monex Force 10,000 Microsilica from WR Grace Co.
3-10
Coarse aggregate: #67 crushed granite Fine aggregate: Natural sand, FM=2.3, natural sand FM=3.0
Manufactured sand FM= 2.9 Set Retarder: WRDA-64 from WR Grace Co. Superplasticizer: Naphthalene based and melamine based high range water reducer,
WRDE 19 and Daracem ML-330 respectively from WR Grace Co. Air entraining admixture: Darex AEA from WR Grace Co.
3.4.1 Mix Designs Initially, Type III cement was used due to the consideration of the high early strength
requirement for this mix. Mix designs using Type III Portland cement alone and with silica fume and fly ash were investigated. These mixes developed in the laboratory were divided into the following five groups:
Group 1: 846 lb/yd3 cement; no mineral admixtures G2-8a and G2-8b
Group 2: 770 to 805 lb/yd3 cement; 13 - 15 % fly ash; no silica fume G2-4 to G2-6
Group 3: 740 to 845 lb/yd3 cement; no fly ash; 74-100 lb. silica fume G2-1 to G2-3
Group 4: 750 lb/yd3 cement; no fly ash; 35 lb. silica fume G2-7, G2-9 to G2-13, and G2-13A.
Group 5: 650 lb/yd3 cement; 100 lb. Fly ash; 35 lb. silica fume. G2-15, G2-15A, G2-18, G2-20, G2-21 and G2-23
Table 3.7 summarizes the mix proportions and the concrete properties for Group 1, Group 2 and Group 3. Table 3.8 and Table 3.9 summarize the mix proportions and the concrete properties for Group 4 and Group 5, respectively.
3-11
Table 3.7 Grade 2 HPC mixes (Group 1, 2 and 3)
Group Number HPC Project Number: Mixing Method:
Cementitious (lb/yd3) Type III Flyash, Class F Silica Fume
Fine Aggregates (lb/yd3) Natural Sand (BB) Manufactured Sand (BC)
Coarse Aggregate (lb/yd3) #67 Crushed Granite (BC)
Water (lb/yd3) Water (Net) (lbs/yd3) Water / Cement Ratio Water / Cementitious Ratio Chemical Admix (oz/cwt) : Set Retarder Superplasticizer (NA) Superplasticizer (ML) Air Entraining (Darex) Plastic Concrete Properties Total Wt. /yd3 (lbs) Unit weight (pcf) Slump (in.) Air Contents (%) Compressive Strength (psi) 24 Hour 3 Day 7 Day 28 Day 56 Day
Group 1
G2-8a G2-8b
Lab
Lab
Group 2
G2 - 5 G2 - 6
Lab
Lab
G2-1 Lab
Group 3 G2-2 Lab
G2-3 Lab
843
844
725
750
845
760
740
92
117
100
76
74
408
409
226
220
750
1067
1014
614
564
550
1759
1763
1718
1689
1799
1724
1695
286
286
262
265
299
250
263
0.34
0.34
0.36
0.35
0.35
0.33
0.36
0.34
0.34
0.32
0.31
0.32
0.30
0.32
0.00
0.00
2.80
2.80
3.00
3.00
3.00
24.98 0.00
29.98 0.00
26.12 1.79
26.10 3.45
29.52 1.62
25.01 2.10
25.01 3.30
3922 145.3 1.75 2.75
3933 145.7 2.50 3.25
3869 143.3 5.25 3.50
3873 143.5 4.75 3.50
3812 141.2 4.75 3.80
3893 144.2 6.50 4.50
3807 141.0 7.00 4.50
6125 6955 7565 8275 8885
6275 7255 8090 9065 9085
6020 7185 7970 9330 9810
6560 7455 8630 9930 10230
6475 8460 9870 11750 12180
7200 9255 11110 13150 13810
6835 8745 9975 12325 N/A
3-12
Table 3.8 Grade 2 HPC mixes (Group 4)
Group Number
HPC Project Number: Mixing Method:
Cementitious (lb/yd3) Type III Flyash, Class F Silica Fume
Fine Aggregates (lb/yd3) Natural Sand (BB) Manufactured Sand (BC)
Coarse Aggregate (lb/yd3) #67 Crushed Granite (BC)
Water (lb/yd3) Water (Net) (lbs/yd3) Water / Cement Ratio Water / Cementitious Ratio
Chemical Admix (oz/yd3) : Set Retarder Superplasticizer (NA) Superplasticizer (ML) Air Entraining (Darex)
Plastic Concrete Properties Unit weight (pcf) Slump (in.) Air Contents (%) Set Time (hr: min)
Compressive Strength (psi) 24 Hour (ambient) (insulated) 3 Day (ambient) (insulated) 7 Day (ambient) (insulated) 28 Day (ambient) (insulated) 56 Days (ambient) (insulated)
CHLORIDE PERMEABILITY 28 Days (coulombs) 56 Days (coulombs)
G2 -7 Lab
G2-9 Lab
G2-10 Lab
Group 4
G2-11 G2-12
Lab
Lab
G2-13 G2-13A G2-14
Lab
Lab
Lab
750 35
419 612
1778
740 35 1037
1754
721 34
0 1010
1710
761 35
0 915
1914
775 36 931
1948
739 35 986
1725
650 100 35
1000
1750
745 35 993
1738
247
261
254
251
255
258
260
260
0.33 0.35 0.35 0.33 0.33 0.35 0.35 0.35
0.31 0.34 0.34 0.32 0.31 0.33 0.33 0.33
23.50
353.00 19.20
23.50
353.00 19.00
23.50
394.00 47.10
23.50
395.00 44.20
23.50
397.00 19.20
23.50
348.00 19.20
23.50
394.00 19.20
23.00 196.00
19.20
143.90 4.25 5.00
142.70 4.00 5.75
139.20 3.75 7.50
144.70 1.75 4.00
147.20 2.25 3.75
139.30 5.00 6.25 6:00
141.60 6.00 7.00
140.20 3.75 5.50 5:40
7360 7340 8455 8160 9305 8850 11175 10640 11460 10780
6430 7335 7965 9340 9695
7610 7990 6536 8605 9225 7909 9080 9525 8886 10465 11170 10377 11045 12175 11145
5985 7513 7210 8110 8100 8335 9830 9641
6798 8190 9453 11189 11717
1080 1005
1140 1205
3-13
Table 3.9 Grade 2 HPC mixes (Group 5)
Group Number
HPC Project Number: Mixing Method: Cementitious (lb/yd3)
Type I Cement Type III Cement Flyash, Class F Silica Fume Fine Aggregate (lb/yd3) Natural Sand Coarse Aggregate (lb/yd3) #67 Crushed Granite Water (lb/yd3) Water (Net) Water / Cement Ratio Water / Cementitious Ratio Chemical Admixtures (oz/yd3) : Set Retarder Superplasticizer (Naphthalene) Superplasticizer (Melamine) Air Entraining
PLASTIC CONCRETE PROPERTY Unit Weight (pcf) Slump (in.) 4-6 in. Air Contents (%) 5-8% Set Time (hr. : min.)
COMPRESSIVE STRENGTH (psi) 24 Hour (Ambient Curing) (Insulated Curing) 3 Day (Ambient Curing) (Insulated Curing) 7 Day (Ambient Curing) (Insulated Curing) 28 Day (Ambient Curing) (Insulated Curing) 56 Day (Ambient Curing) (Insulated Curing)
CHLORIDE PERMEABILITY 28 Days (coulombs) 56 Days (coulombs)
G2-15 G2-15A
Lab
Lab
644
658
99
101
35
36
991
1013
1734
1772
253.00 0.39 0.33
258.00 0.39 0.33
23.00
347.00 13
23.85 199.00
13.5
139.90 6.50 5.50 6:48
142.80 7.00 5.00 6:00
5580 7588 8304 10422 11091
6187 8054 7985 9233 9469 9800 11250 10990
12100
871 701
Group 5
G2-18 Truck
G2-20 Lab
650
650
100
100
35
35
1040
1000
1710
1750
208
208
0.32
0.32
0.26
0.26
21.00
220.00 16.00
143.4 5.50 5.30
21.00
487.00 16.00
141.8 4.00 4.80
--6148 5852 7237 7297 7522 8535 8299 N/A N/A
6090 7850 8258 9044 9780 9758 11434 11513 13206 12316
G2-21 Lab
650 100 35
1000
1750
235 0.36 0.30
21.00
338.00 16.00
141.0 4.00 5.00
5806 7171 7719 7771 8692 8999 10978 10711 12225 12001
G2-23 Lab
650
100 35
1000
1750
215 0.33 0.27
21.00
510.00 16.00
143.5 3.00 4.50
6730 9793 8637 10349 10460 10419 11960 11524 12682 N/A
3-14
After considering all the laboratory mix designs, it was decided to use G2-20 (same as G215) to evaluate the mechanical and durability properties for Grade 2 HPC. This mix used an extra 25 lb/yd3 of cement (675 lb. instead of 650 lb.) and Type I cement was used. This mix was designated as G2-27. The final mix design, after adjustment in the plant batching operation, is shown in Table 3.10
Table 3.10 Precast Concrete Plant Mix G2-27
HPC Project Number: Mixing Method: Cementitious (lb/yd3)
Type I Cement Type III Cement Flyash, Class F Silica Fume Fine Aggregate (lb/yd3)
Natural Sand Coarse Aggregate (lb/yd3)
#67 Crushed Granite Water (lb/yd3)
Water (Net) Water / Cement Ratio Water / Cementitious Ratio Chemical Admixtures (oz/yd3) : Set Retarder Superplasticizer (Naphthalene) Superplasticizer (Melamine) Air Entraining
PLASTIC CONCRETE PROPERTY Unit Weight (pcf) Slump (in.) 4-6 in. Air Contents (%) 5-8%
COMPRESSIVE STRENGTH (psi) 24 Hour (Ambient Curing) (Insulated Curing) 3 Day (Ambient Curing) (Insulated Curing) 7 Day (Ambient Curing) (Insulated Curing) 28 Day (Ambient Curing) (Insulated Curing) 56 Day (Ambient Curing) (Insulated Curing)
Chloride Permeability (AASHTO T-277) 56 Day (coulombs)
G2-27 Plant
675
100 33
1000
1750
208 0.31 0.26
21.00
188.00 16.00
144.1 7.00 5.00
6161 8351 7751 9196 8669 9264 10494 10152 11619 11284
726
3-15
3.4.2 Curing Temperature Study A total of four different curing methods were used for the curing temperature study.
Three of these methods were for cast cylinders, and a fourth method was for cylinder cores taken from the prototype beam member. All specimens were 4 in. diameter by 8 in. long cylinders.
ASTM Curing: ASTM curing consisted of casting and covering the 4 x 8 in. cylinders with plastic lids to prevent moisture losses. The cylinders were maintained in the ambient conditions at the time of casting. A thermocouple was inserted into one of the cylinders for 24 hours to monitor the temperature profile. After the 24 hour period, the plastic molds were removed and the cylinders were cured in lime saturated water or in a fog room at 100% humidity and at 73 oF until they were ready to be tested.
Insulated Curing: Insulated curing consisted of casting and covering 4 x 8 inch cylinders with plastic lids to prevent loss of moisture. The cylinders were then placed inside an insulated box, a thermocouple was inserted in one cylinder, and the temperature was monitored for 24 hours. The box was constructed out of -inch plywood boards, and the box measured 15 inches high, 20 inches wide, and 30 inches long. The inside of the box was lined with 2 inch thick foiled-backed rigid insulation sheets. The purpose of using the insulation box was to simulate the insulated conditions that occur in the middle of a concrete girder. Upon completion of the 24 hour curing in the insulated box, the cylinder specimens were stripped from the plastic molds and cured in lime saturated water or in a fog room at 100% humidity and at 73 oF until the required testing age.
Temperature Match Curing (Sure-Cure): The equipment for the temperature match cure procedure consisted of a temperature sensing and heating control box, thermocouple wires and the temperature regulated 4x8-in. cylinder specimen molds. The control box controlled the temperatures of concrete cylinders allowing the cylinder temperature to follow the time temperature profile of an actual curing concrete member. Thermocouples were inserted into a prototype concrete beam, see Figure 3.3, as well as the cylinder mold setup. The temperature in the concrete beam was monitored and the temperature profile was used to regulate and maintain the temperatures in the match cure cylinder specimen to the same temperature as that of the concrete beam. Thermocouples in the beam and temperature control box showed that the temperatures of the beam and the cylinder were within a 2 oF difference.
3-16
Beam Temperature Study: This study was conducted to evaluate the viability of using the insulated curing method to model the concrete temperature of an actual concrete girder. A prototype section of the lower half of an AASHTO type IV beam was constructed and used to obtain cores for comparison with the concrete strength from the match cured and insulated cured procedures. Thermocouples were embedded in the middle of the beam, and the temperature in the beam was used to control the temperature of the match cure cylinder specimens. Ambient and insulated cured cylinders were cast as well. Figure 3.3 shows the dimensions of the prototype beam section and the locations of the thermocouples placed in the beam. 4 in. diameter by 26 in. long cores were taken from the bottom flange of the beam at different curing ages.
Three mixes, G2-13B, G2-15B and G2-13C were produced. Strength gain curves with the different curing conditions are shown in Figures 3.4 through 3.6. Temperature curves are given in Figures 3.7 and 3.8.
The temperature effect studies from these three grade 2 HPC mixes produced many significant results. The maximum concrete temperature in the beams and in the 14 by 36 by 30 in. block was significantly higher (as much as 75 oF) than the concrete in the ASTM cured specimen during the initial curing period. The concrete temperature in the match cured specimens was about 2 to 5 oF higher than that in the beam. The concrete temperature in the insulated cured specimens was comparable with that of the beam. The 24-hour strength of the concrete in the beam was significantly higher than that of the ASTM cured specimens. The concrete strengths from the match cured specimens and the insulated cured specimens were comparable and both of them were comparable to the concrete strength in the beams (maximum difference of about 8%). The results of this study demonstrated that the compressive strength of the concrete specimens under the insulated box curing method matched closer with the strengths from core specimens taken from the full-size beam than that under the standard ASTM curing.
3-17
Figure 3.3 Prototype AASHTO Type IV beam for temperature curing study. 3-18
Compressive Strength, psi
12000 11000 10000
9000 8000 7000 6000
0
ASTM Curing Insulated Beam core (C) Beam core (end)
7 14 21 28 35 42 49 56 Days
Figure 3.4 Strength gain for different curing conditions (G2-13B)
Strength, psi
13000 12000 11000 10000
9000 8000 7000 6000
0
ASTM Curing Ins ulated March Cure Beam core (C) Beam core (end)
7 14 21 28 35 42 49 56 Days
Figure 3.5 Strength gain for different curing conditions (G2-15B)
3-19
Strength, psi
11000
10000
9000 8000 7000 6000 5000
0
ASTM Curing Ins ulated March Cure Beam core Block core
7 14 21Day2s8 35 42 49 56
Figure 3.6 Strength gain for different curing conditions (G2-13C)
G2-15B HPC Temperature History Under Different Curing Conditions 160
140
120
Temperature, F
100
80
60
40
0
2
4
6
8 Time1,0hour 12
14
16
18
20
Room Temp match cure
room beam, center
insulated beam, side
Figure 3.7 Temperature profiles under different curing conditions (G2-15B)
3-20
180
160
140
Temperature, F
120
100
80
60
0
4
Room Temp beam CF
8
12
16
Hour
room beam, side
insulated block
20
24
beam center
Figure 3.8 Temperature profiles under different curing conditions (G2-13C)
3.4.3 Grade 2 Properties The main property study was done on the plant mixed G2-27 mix. Compressive
strength results are shown in Table 3.11 and Figure 3.9 Table 3.11 Compressive strength of G2-27
Age
1 day 3 day 7 day 28 day 56 day
Compressive Strength, psi
Ratio of
Insulated
ASTM Curing
4in./6in. cyl
4-in. dia cyl 4-in. dia cyl 6-in. dia cyl ASTM
8351
6016
9196
7751
7491
1.03
9264
8670
8267
1.05
10152
10494
9936
1.06
11284
11619
10848
1.07
Average Difference =
1.05
3-21
Compressive Strength, psi
12000 11000 10000
9000 8000 7000 6000
0
4x8 INS 4x8 ASTM 6x12 ASTM
7 14 21 28 35 42 49 56 Days
Figure 3.9 Compressive strength of G2-27
The modulus of rupture at 56 days for a 4x4x14 in. beam averaged 832 psi. The 56-day modulus of elasticity was 4.87 ksi with a Poisson's ratio of 0.17. The average rapid chloride permeability at 56-days was 330 coulombs.
To ensure that Grade 2 HPC can be produced statewide using locally available aggregates, two #67 coarse aggregates from Southern Aggregate's Postell Quarry and Martin Marietta's Dan Quarry in addition to the coarse aggregate from the Blue Circle's Lithonia Quarry were used in this study. Three Grade 2 HPC mixes using the identical mix design were produced using the coarse aggregate from these three sources. These three mixes were designated as G2-29, G2-30, and G2-31 as presented in Table 3.12. The general conclusion is that aggregates from these three sources can be used for the production of Grade 2 HPC.
It was concluded that Grade 2 HPC mixes for precast prestressed concrete bridge girders can be produced utilizing the conventional constituent materials, conventional production and curing procedures to meet the following strength and the durability requirements. Both the naphthalene based HRWR and the melamine based HRWR are effective in improving the workability of the Grade 2 HPC mixes. The concrete strengths from the match cured specimens and the insulated cured specimens were comparable, and both of them were comparable to the concrete strength in the beams. The strengths from ASTM cured cylinders at 24-hours was significantly lower. The average rapid chloride ion penetration of the concrete specimens tested at 56-day age was 330 coulombs, which was significantly lower than the requirement of 2000 to 3000 coulombs, and represents very low permeability.
3-22
Finally, it was concluded that Grade 2 HPC can be produced using aggregates from all regions of Georgia.
Table 3.12 Mix proportions of Grade 2 HPC using three statewide aggregate sources.
HPC Project Number:
Aggregate Source
Quantity CONSTITUENT INGREDIENTS Cementitious: (lbs)
Type III Flyash, Class F Silica Fume Fine Aggregate: Natural Sand (lbs) Coarse Agg. #67 Granite (lbs)
Water: (Net) (lbs) Water / Cement Ratio Water / Cementitious Ratio
Chemical Admixture (oz) : Set Retarder Superplasticizer (Naphthalene) Air Entraining (Darex)
PLASTIC CONCRETE PROPERTIES Unit weight (pcf) Slump (in.) Air Contents (%)
COMPRESSIVE STRENGTH (psi) 24 Hour 3 Day 7 Day 28 Day 56 Day
G2 1.0 yd3
G2-#29 Blue Circle
1.1 ft3
G2-#30
Augusta 1.1 ft3
G2-#31
Macon 1.1 ft3
650 100 35 1000 1750 215 0.33 0.27
21 16 315
26.5 4.1 1.4 40.7 71.3 8.8 0.33 0.28
0.91 16.2 0.71
147.6 1 4.5
7291 7689 8818 10359 12461
26.5 4.1 1.4 40.7 71.3 8.8 0.33 0.28
0.91 16.2 0.71
148 1 4.7
7066 7256 8707 10753 10566
26.5 4.1 1.4 40.7 71.3 8.8 0.33 0.28
0.91 14.5 0.71
148.8 1 4.3
7961 8573 9660 11459 11677
3-23
3.5 Grade 4 HPC The goal of Grade 4 HPC was to develop a concrete with a minimum specified design
strength of 14,000 psi and an average mix strength of 15,400 psi at 56 day-age for precast prestressed concrete bridge girders. Also required was chloride permeability between 2,000 to 3,000 coulombs. No air content was specified, but the freeze-thaw durability (ASTM C666) should be good. Further, the slump should be between 4 and 6 inches.
The materials used for developing the Grade 4 HPC mixes in the laboratory are summarized below:
Cement:
Type I and Type III cement from Blue Circle Atlanta Plant
Fly ash:
Type F fly ash from Monex
Silica Fume: Force 10,000 microsilica from WR Grace Co.
Coarse aggregate: #7 crushed granite and #67 crushed granite
Fine aggregate: Natural sand, FM=2.3, natural sand FM=3.0
Manufactured sand FM= 2.9
Set Retarder: WRDA-64 from WR Grace Co.
Superplasticizer: Naphthalene based and melamine based high range water reducer,
WRDA-19 and Daracem ML-330, respectively, from WR Grace Co.
Initially, Type III cement was used due to the consideration of the high early strength requirement for this mix. Mix proportions using Type III Portland cement alone and with silica fume and fly ash were investigated. These mixes were divided into the following four groups:
Group 1: 1000 lb/yd3 cement; 200 lb/yd3 silica fume, no fly ash: G4-1, and G4-4 to -10
Group 2: 800 lb/yd3 cement; 100 lb/yd3 fly ash; 100 lb/yd3 silica fume: G4-2, G4-3, G4-8, G4-11 to G4-14 Group 3: 800 lb/yd3 cement; 100 lb/yd3 fly ash; 60 lb/yd3 silica fume: G4-15 and G4-16 Group 4: 800 lb/yd3 cement; 100 lb/yd3 fly ash; 80 lb/yd3 silica fume: G4-17 to G4-19 Tables 3.13 through 3.15 summarize the mixes and properties.
3-24
Table 3.13 Grade 4 HPC Mixes (Group 1)
Group Number HPC Project Number:
Mixing Method: (1)
G4-1 TD
G4-4 TD
G4-5 TD
Group 1
G4-6 TD
G4-7 TD
G4-9 BM
Cementitious (lb/yd3)
Type I
Type III
934
950
950
939
939
964
Flyash, Class F
Silica Fume Fine Aggregates (lb/yd3)
187
190
190
188
188
193
Natural Sand (MM)
845
Natural Sand (BB)
233
233
226
220
750
Manufactured Sand (BC) Coarse Aggregate (lb/yd3)
582
582
564
550
#7 Crushed Granite (BC)
1595
1718
1689
1799
#7 Florida Rock Limestone Water (lb/yd3)
1572
1595
Water (Net) (lbs/yd3)
326
296
296
262
265
299
Water / Cement Ratio
0.35
0.31
0.31
0.36
0.35
0.35
Water / Cementitious Ratio Chemical Admix (oz/yd3) :
0.29
0.26
0.26
0.32
0.31
0.32
Set Retarder
53.00 57.00 57.00 31.00 31.00 35.00
HRWR (NA) (2)
741.00 741.00 750.00 750.00 777.00
HRWR (ML) (2) Air Entraining Admix. Plastic Concrete Properties
549.00
N/A
N/A
N/A
N/A
N/A
N/A
Unit weight (pcf)
144.80 144.65 144.65 143.3 143.5 141.2
Slump (in.)
7.00
8.50
8.50
5.25
4.75
4.75
Air Contents (%)
N/A
N/A
N/A
3.50
3.50
3.80
Compressive Strength (psi) (3)
24 Hour
6670
2255
1775
6020
6560
6475
3 Day
8080
9020
8080
7185
7455
8460
7 Day
9,265 10,365 9,140
7970
8630
9870
28 Day
11,435 12,470 11,480 9330
9930 11750
56 Day
12,640 13,320 12,280 9810 10230 12180
(1) TD = Tilt Drum Mixing; BM = Blade Mixing
(2) NA = naphthalene based HRWR, WRDA-19; ML = melamine based HRWR, Daracem ML-330 (3) Under ASTM cured conditions
G4-10 BM
964 193
1067
1724
250 0.33 0.30
35.00 777.00
N/A 144.2 6.50 4.50 7200 9255 11110 13150 13810
3-25
3-26
Table 3.15 Grade 4 HPC Mixes (Groups 3 and 4)
Group Number
Group 3
HPC Project Number: Mixing Method: (1)
Cementitious (lb/yd3)
G4-15 BM
G4-16 BM
Type I
840
Type III
838
Flyash, Class F
96
99
Silica Fume Fine Aggregates (lb/yd3)
59
59
Natural Sand (BB) Coarse Aggregate (lb/yd3)
907
909
#7 Crushed Granite (BC)
1815
1818
#67 Crushed Granite (BC) Water (lb/yd)
Water (Net) (lbs/yd3)
256
256
Water / Cementitious Ratio Chemical Admix (oz/yd3) :
0.26
0.26
Set Retarder
28.00
28.00
HRWR (NA)
428.00 335.00
HRWR (ML) Plastic Concrete Properties
Unit Weight (pcf)
148.20 148.50
Slump (in.)
6.00
5.50
Air Contents (%) Compressive Strength (psi)
N/A
N/A
24 Hour (Ambient Curing)
7785
7140
(Insulated Curing)
12333 11052
3 Day (Ambient Curing)
9169
9161
(Insulated Curing)
12761 11764
7 Day (Ambient Curing)
11270 10832
(Insulated Curing)
13142 11764
28 Day (Ambient Curing)
13430 13174
(Insulated Curing)
13409 12953
56 Day (Ambient Curing)
13450 13995
(Insulated Curing)
14151 13526
(1) BM = Blade Mixing TD = Tilt Drum Mixing
G4-17 BM
Group 4 G4-18 BM
G4-19 BM
835
835
835
98
98
100
79
79
80
904
904
900
1808
1808
1800
256
256
256
0.25
0.25
0.25
28.00 392.00
28.00 392.00
28.00 432.00
148.50 6.25 N/A
148.50 7.00 N/A
148.80 1.50 N/A
7335 12207 9443 12806 11386 13156 13790 13418 15360 14719
7467 11502 9357 12305 11339 12612 13673 13178 14465 13780
9316 13031 11401 13525 13229 13825 14218 14497
NA NA
After considering all the mix designs developed in the laboratory, it was decided to use G4-19 to evaluate the mechanical and durability properties for Grade 4 HPC made at the precast concrete plant. Three batches of Grade 4 HPC mixes (G4-21, G4-22 and G4-23) were produced for this purpose. The mix proportions of these three mixes are identical to the G4-19 mix, except the amount of the superplasticizer used for these mixes was substantially less than that for the G4-19 mix. In the first two batches, water content of the aggregate was not
3-27
controlled, resulting in weak mixes. Moisture content of aggregate was measured in G4-23, so the exact water content was controlled. The mixes are shown in Table 3.16 while Figure 3.10 shows the compressive strength gain over time.
Table 3.16 Grade 4 HPC Precast Plant Mixes
HPC Project Number:
Constituent Ingredients:
Cement: Type I
Type III
Flyash, Class F
Silica Fume
Fine Aggregates:
Natural Sand FM =2.4
Coarse Aggregate:
#67 Crushed Granite
#7 Crushed Granite
Water (Net)
W/Cementitious Ratio
Set Retarder
(oz/yd)
Superplasticizer (ML), (oz/yd)
Superplasticizer (NA), (oz/yd) Plastic Concrete Properties:
Unit weight (pcf)
Slump (in.)
Set Time (hr:min) Compressive Strength:
24 Hr. (ambient curing)
(insulated curing)
3 Day (ambient curing)
(insulated curing)
7 Day (ambient curing)
(insulated curing)
28 Day (ambient curing)
(insulated curing)
56 Day (ambient curing)
(insulated curing)
Chloride permeability
56 Day (coulomb)
G4-19 Lab/BM
835
100 80
900
1800
256 0.25 27.00
447.00
148.80 1.50
G4-21 Plant
835
100 80
900
1800
245 0.24 27.00
138.00
149.10 5.00
G4-22 Plant
835
100 80
900
1800
256 0.25 27.00
274.00
149.50 7.00
G4-23 Plant
835
100 80
900
1800
245 0.24 27.00
306.00
152.90 5.00
9316
8368
8600
13031
10205
11240
13400
11401
6692
10138
10200
13525
10365
11909
14300
13229
7326
11600
12200
13825
10367
12400
13700
14218
10115
13213
14840
14497
10739
13390
15385
10403
13681
16203
10734
13728
15761
71
3-28
18000
16000
Compressive Strength, ps
14000
12000 10000 8000 6000
0
G4-21-INS G4-21-ASTM G4-22-INS G4-22-ASTM G4-23-INS G4-23-ASTM
7 14 21 28 35 42 49 56 Days
Figure 3.10 Compressive strength gain with age for G4 precast plant mixes
The G4 modulus of rupture using 4x4x14 in. beams was 795 psi at 56 days; the modulus of elasticity at one day was 4.73 ksi and at 56 days was 4.78 ksi. The 56-day rapid chloride permeability test resulted in 71 coulombs.
This study was initiated to determine if Grade 4 HPC design could be produced using aggregates from three different sources. The sources of the aggregates were the same as those used in the Grade 1 and Grade 2 analyses. By the 56-day age, the mixes had achieved 17,005 psi, 15,676 psi and 16,687 psi for the Blue Circle, Augusta and Macon aggregates respectively. The strengths obtained surpassed the 15,400 psi required average compressive strength by the 56-day testing age. Based on the results obtained, it was concluded that the Grade 4 HPC can be produced using these three aggregate sources from across Georgia.
It was concluded that Grade 4 HPC mixes for precast prestressed concrete bridge girders could be produced utilizing the conventional constituent materials, conventional production and curing procedures to meet the following strength and the durability requirements. The Grade 4 HPC mixes could be produced in the laboratory with the blade mixer and by following the procedures specified by the ASTM C 192. With the extremely high
3-29
cementitious content and low water content in the Grade 4 HPC mixes, the use of tilt-drum mixer became ineffective. For precast plant production of Grade 4 HPC, it was important to accurately determine the moisture content of the aggregates and to adjust the mix proportions accordingly.
3.6 Creep and Shrinkage Properties of HPC
Precast prestressed beams and girders made using high strength, high performance concrete (HPC) may reach their release strength at an early age (i.e. less than 24 hours), and they may be subjected to the high, sustained prestress loading at that early age. As shown in the previous temperature study, the accelerated curing affects the early strength gain. Creep and shrinkage data are essential for understanding and for predicting the deformational history of such pretensioned members. The bulk of published data on creep and shrinkage of HPC were for specimens which either were loaded at ages greater than 24 hours or for specimens which were allowed to cure under ambient conditions according to ASTM standards. Also current prediction models were either developed using normal strength concrete or calibrated using the limited data available on HPC.
The objectives of this part of the research were to determine the creep and shrinkage characteristics of HPC loaded as early as 12 hours after casting, and to develop a prediction model for creep and shrinkage deformations which accounts for the properties of HPC under both early- and late-age loading. To account for the effect of the initial curing temperature, test specimens were accelerated-cured to emulate the curing temperatures developed in full-scale members. However, standard-cured specimens also were tested to correlate the results of this study with the data reported in the literature.
Concrete cylinders measuring 4 in. x 15 in. were monitored for more than a year to measure their creep and shrinkage strains. The mixes were designated as grade 2 (G2) and grade 4 (G4) and were identical to mix G2-21 (Table 3.9) and to mix G4-19 (Table 3.16). A total of nineteen creep and shrinkage tests were initiated at ages 12 hrs, 16 hrs, 20 hrs, 24 hrs, 3 days, 7 days, and 28 days. The applied stress ratios varied between 16% and 60% of the ultimate compressive strength at the time of loading. Each test included two concrete cylinders that were loaded in a vertical creep frame (shown in Figure 3.11) and a third cylinder, which
3-30
was left unloaded for shrinkage and thermal strain measurements. Each cylinder contained two sets of diametrically opposite embedded gage points 10 in. apart; these strain gauge deformations were measured using a detachable mechanical (DEMEC) gage. During the test period, the cylinders were stored in an environmental chamber at 50% 4% relative humidity and 23C 2C (73F 3.5F) temperature.
Figure 3.11 Creep frames with 4x15 inch cylinder specimens
Two initial curing methods were used in preparing the test cylinders: the standard curing method according to ASTM C 512 and accelerated curing method in an insulated box.
The time-dependent strain data and the test parameters for all the tests conducted in this study are summarized in Table 3.17. The creep strains were calculated by subtracting from the total strains measured on the loaded cylinders the shrinkage and the thermal strains measured on the unloaded cylinder. The creep data are presented in Table 3.17 in terms of total creep strain, creep coefficient (ratio of creep strain to initial strain) and specific creep (creep strain per unit applied stress). The creep coefficient is the most common format by which creep is included in concrete structural analysis; hence, it is emphasized herein.
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Table 3.17 Time-dependent strain data
Test Grade 2
Initial Age at f'co, curing loading psi
Shrinkage
Specific
f'c, / f'co, Test
Total
Initial
Creep strain, Creep creep,
psi
% period, days strain,10-6 strain, 10-6 strain, 10-6 10-6 coefficient 10-6/psi
G2-27
Insulated 24 h 8350 10,150 60
461
Standard 3 d 7610 10,490 60
459
G2-32
Insulated 2 d 12,380 12,940 16
403
Insulated 2 d
403
Standard 28 d* 14,140 14,140 50
376
G2-33
Insulated 12 h 3100 12,750 40
360
Insulated 16 h 6530
360
Insulated 20 h 8350
360
Insulated 24 h 9005
360
Insulated 7 d 11,550
354
Grade 4
3244 3296 1017 951 2406 1841 2287 2598 2580 2724
1209 1227 413 397 1250 515 859 1015 1089 1152
1516 1561 283 284 939 540 676 840 883 1151
520
1.254 0.318
509
1.272 0.327
321
0.684 0.059
270
0.716 0.06
218
0.752 0.197
594
1.05 0.424
589
0.786 0.257
587
0.827 0.251
548
0.811 0.247
421
0.999 0.249
G4-23
Standard 3 d 10,350 14,840 60
446
G4-25
Insulated 2 d 14,400 15,520 19
403
3571
1339
1672
560
1187
506
346
335
Insulated 2 d
403
1205
495
330
380
Standard 28 d* 16,380 16,380 50
376
G4-26
Insulated 12 h 5520 14,775 40
373
2554
1361
939
254
2792
952
1070
548
Insulated 16 h 10,075
373
2780
1144
1029
407
Insulated 20 h 10,525
373
2806
1193
977
447
Insulated 24 h 12,040
373
2883
1282
992
439
Insulated 7 d 11,970
367
2969
1229
1246
492
1000 psi = 6.895 MPa.
Each strain value is the average of two cylinders except for shrinkage, which is for one cylinder.
Each compressive strength result is the average of three 4-in. x 8-in. (100-mm x 200-mm) cylinders cured identically to the creep cylinders.
All the long-term strain data and their derivatives are calculated at the end of the test period. f'co & f'c = compressive strength at the loading age (to) and at 28 days, respectively.
h = hours & d = days The test ID reflects the HPC concrete grade and a laboratory batch number.
Mixes batched at a precast concrete plant in 1.5- to 2.0-yd3 (1.15- to1.53-m3) volumes
Laboratory mixes which were prepared in 1.1-ft3 (0.03-m3) volumes using a fixed-drum mixer.
* The shrinkage strains for this test were estimated from the results of the 2-day tests for the same mix using
the procedure outlined in the ACI Committee 209 report.
1.249 0.684 0.667 0.69 1.124 0.899 0.819 0.774 1.014
0.35 0.072 0.069 0.197 0.48 0.254 0.232 0.204 0.261
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The creep and shrinkage data were analyzed using a nonlinear regression analysis to obtain prediction formulas for HPC. The creep coefficient data were curve-fitted to the general hyperbolic equation suggested by ACI Committee 209, Equation 3-1.
(t, to )
=
d
(t - to ) + (t - to )
u
(3-1)
where (t,to) is the creep coefficient at age t (in days) for a concrete loaded at age to (in days), d (in days) and are constants, and u is the ultimate value of the creep coefficient. The constant was set equal to 0.6 as recommended by ACI Committee 209 for normal strength concrete because no significant improvement in the curve-fits was obtained by treating it as a variable.
The shrinkage data were curve-fitted to the following equation:
sh (t, to )
=
f
(t-to ) +(t -to
)
(
sh
)
u
(3-2)
where sh(t, to) is the shrinkage strain at age t (in days) for a concrete started to dry at age to (in days), and f (in days) are constants, and (sh)u is the ultimate shrinkage strain. The value of the exponent was set equal to 0.5 because treating it as a variable did not improve the accuracy of the curve fits. The time function in Eq. (3-2), which is used in the simplified Bazant and Panula (1980) model and in the European code model (CEB-FIP), was selected instead of that recommended by ACI Committee 209 because the former was found to better describe the high initial drying rate of HPC. The results of the regression analyses of both the creep and the shrinkage data are summarized in Table 3.18.
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Test Grade 2
HPC G2-27 G2-32
G2-33
Grade 4 HPC
G4-23 G4-25
G4-26
Table 3.18 Results of the Regression Analysis
Age at
loading
d, days
Creep u
Shrinkage
R2
f, days
(sh)u x 10-6
24 h
7.14
1.42 0.987 71.1
536
3 d
4.55
1.35 0.987 34.8
507
2 d
8.18
0.86 0.970 25.1
409
2 d
9.00
0.83 0.910 18.9
384
28 d
6.64
0.91 0.977
--
--
12 h
1.01
1.02 0.940 17.4
559
16 h
1.72
0.86 0.986 23.1
551
20 h
2.45
0.88 0.992 15.8
540
24 h
1.52
0.80 0.973 15.9
494
7 d
2.69
1.04 0.991 14.3
396
3 d
1.99
1.25 0.996 29.7
569
2 d
5.12
0.76 0.975 29.8
464
2 d
6.12
0.82 0.954 36.6
422
28 d
7.57
0.84 0.969
--
--
12 h
1.94
1.17 0.986 15.3
529
16 h
3.40
0.91 0.987 22.7
422
20 h
3.81
0.85 0.988 25.5
415
24 h
3.31
0.76 0.988 20.9
398
7 d
3.51
1.07 0.995 18.4
468
R2
0.961 0.951 0.972 0.923
-- 0.950 0.940 0.944 0.962 0.955
0.981 0.969 0.973
-- 0.970 0.969 0.961 0.947 0.965
Effect of Concrete Compressive Strength on Creep The data for the cylinders that were loaded at ages between one and two days with a
maximum stress/initial strength ratio of 0.4 indicate that the creep coefficient decreased as the concrete compressive strength increased. This trend is attributed to the dense cement paste matrix (the seat of creep in concrete) and the low water-cementitious materials (w/cm) ratio which corresponds to high strength. The low volume of voids present in a dense paste permits limited consolidation of the cement gel under pressure which results in reduced creep. Further, the low water content in the paste may influence creep in two ways: that of reducing the
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quantity of evaporable water and that of reducing the amount of interlayer water held between the gel particles. The former would reduce drying creep, while the latter would hinder the relative movement of gel particles under pressure, hence, reducing basic creep.
The regression analysis suggests that the ultimate creep coefficient for HPC having a 28-day compressive strength between 13,000 and 15,500 psi, an age at loading of one to two days, and a maximum stress/initial strength ratio of 0.4 is, on average, between 0.83 and 0.78, respectively. These values are, respectively, only 35% and 33% of the average value recommended by ACI Committee 209 for normal strength concrete (NSC).
Effect of concrete age at loading on creep
HPC appears to behave similar to NSC with respect to the concrete age at loading, that is the older the concrete at the time of loading, the less creep deformation it exhibits. This trend can be seen in Figure 3.12, which presents measured creep coefficients for test series G2-33 and G4-26 and compares them to the prediction of the ACI Committee 209 model for a loading age of 24 hours. This trend may be explained in terms of the degree of hydration of the cement paste. As concrete ages, the hydration process develops stronger bonds between the cement particles, which, in turn, hinder their relative movements under stress. Also, the progress of the hydration process results in more hydration products which fill some of the voids in the cement paste leaving less room for compaction under load, i.e. less creep.
As shown in Figure 3.12, in addition to exhibiting significantly less creep than NSC as represented by the ACI-209 curve, HPC showed a high initial creep rate. The high initial creep results from the very low porosity of HPC. Water is held firmly within the cement paste matrix; therefore, only water near the surface is allowed to diffuse to the surrounding medium permitting an initially rapid compaction of the paste. Further creep is much slower since it is driven by the diffusion of much more firmly held, interlayer water molecules. Hence, the majority of creep in HPC takes place within the first two weeks causing the seemingly high initial creep rate.
Effect of applied stress/strength ratio on creep
Except for a concrete loaded at the age of one to three days, ACI Committee 209 regards creep of NSC to be linearly related to the applied stress/strength ratio up to a ratio of
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0.4. In this study, measured creep strains for insulated-cured, HPC cylinders loaded at the age of one to two days were found to be linearly proportional to the applied stress/strength ratio up to a ratio of 0.6. The calculated ultimate creep values for the same tests were found to be independent of the stress ratio up to a ratio of 0.4, but directly proportional to the stress ratio for values greater than 0.4. Accordingly, the ACI Committee 2099 type creep coefficient model may be used with a correction factor to account for the increase in the ultimate creep coefficient when the concrete is sustained-loaded with stress/strength ratio between 0.4 and 0.6.
Creep coefficient .
2 1.5
1 0.5
0 0
ACI-2099
Age at loading (hrs): 24
12
16 20
24
f' c = 14,775 psi (102 MPa)
//fc'i'co==0.04.4
100
200
300
400
Time after loading, days
Figure 3.12 Creep coefficient for HPC, series G4-26
Effect of concrete compressive strength on drying shrinkage
For insulated-cured cylinders having a compressive strength between 10,150 psi and 15,520 psi (70 MPa and 107 MPa) and an age at the onset of drying between 24 to 48 hours, measured shrinkage strains after approximately one year varied between 270 microstrains () and 548 . These values are well within the values reported in the literature for HPC. However, the scatter in the measured data and the contradictory findings reported by
3-36
other investigators suggest that, for the time being, shrinkage of HPC may be considered independent of the compressive strength. It should not be overlooked, however, that HPC does exhibit less shrinkage strains than those of NSC drying under the same conditions. The measured shrinkage strains reported above are approximately 40% and 81% of the average value recommended by ACI Committee 209 for NSC.
Effect of drying age on shrinkage
Figure 3.13 presents measured shrinkage strains for test series G4-26 and compares the data with the average value recommended by ACI Committee 209. Except for the 16-hour test, the data in Figure 3.13 suggest that the older the concrete at the beginning of drying, the less it shrinks. A similar trend also was observed for test series G2-32. This behavior may be explained in terms of the degree of hydration of the cement paste at the start of drying. As concrete matures, more cement particles are hydrated leaving fewer water molecules for evaporation (i.e., less drying shrinkage) and fewer cement particles for further hydration (i.e., less autogenous shrinkage).
Figure 3.13 also suggests that HPC exhibits more shrinkage than that of NSC during the first few weeks of drying, but the opposite is true at later ages. This may be explained as follows. At early ages, owing to its high cement content, HPC experiences comparable amounts of drying shrinkage, but more autogenous shrinkage than that of NSC. However, at later ages and as the hydration process develops, autogenous shrinkage becomes increasingly small for both concretes, and drying shrinkage contributes the most to the total shrinkage. As such, owing to its low porosity, HPC exhibits less drying shrinkage and hence less overall long-term shrinkage than that of NSC.
Prediction of Creep and Shrinkage
The creep and shrinkage data obtained in this study were compared to predictions of four current models: the ACI Committee 209 model, the simplified Bazant and Panula (1980) model (BP2 model), the CEB-FIP (1991) model, and the AASHTO LRFD (1998) model. The evaluation of how well a model predicted the data was done by calculating the coefficient of variation as described by Bazant and Panula (1979), with the exception of using the curve fits instead of hand-smoothed curves to describe the data. The curve fits were used because they described the data well, evident from the high values of R2 in Table 3.18. The actual data points
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were not used in the comparison to minimize the effect of random measurement errors or random environmental changes associated with certain data points. The sampling points were selected to be equally spaced on a logarithmic scale to avoid any bias towards early-age or lateage responses. Four sampling points per decade with a maximum value of 10,000 days were used in the comparison.
Shrinkage strains, 1 -06 .
800 700 600 500 400 300 200 100
0 0
ACI 2099
Age at onset of drying (hrs): 24
12 20 24
16
f' c = 14,775 psi (102 MPa)
100
200
300
400
Time after onset of drying, days
Figure 3.13 Effect of early age drying on shrinkage
For each data set, j, the variance of the formulas of each model from the curve fit was calculated as:
S j =
1 n -1
n i =1
2ij
(3-3)
where i = 1, 2 ...n are the sampling points and ij are the deviation of the formulas from the curve fit at each sampling point. The coefficient of variation for a data set, j, was determined by the following equation:
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j
=
Sj J jm
with
J jm
=
1 n
n i =1
J ij
(3-4)
where Jij are the values of the curve fit at the sampling points and Jjm is their mean. Finally, the overall coefficient of variation, , for all the data sets was determined as:
=
1 N
N
2j
j =1
(3-5)
where N is the total number of data sets. Figures 3.14 and 3.15 compare the measured creep and shrinkage data, respectively, for the 24-hour test of series G2-27 with the predictions of the analytical models. Similar results were found for each test series.
With respect to creep, both the statistical data and the graphical comparison indicate that, in general, all the models overestimated the measured creep data. Of the four models, the AASHTO LRFD model gave the closest predictions. The ACI 209, the CEB-FIP and the BP2
models were based on empirical data using NSC data (fc ' less than 6000 psi). Furthermore, the
modifications to the BP2 model to include HPC were based on short-term data (90 days) of
concrete with fc' less than 9,800 psi (Bazant and Panula 1984).
The relatively improved predictions of the AASHTO LRFD model, which is very similar to the ACI 209 model, are the result of incorporating a strength correction factor based on recent studies of creep of HPC.
With respect to shrinkage, the example in Figure 3.15, as with all measured data sets, showed that both ACI 209 and AASHTO models overestimated the test data, while both BP2 and CEB-FIP models underestimated them. According to the results of the statistical analysis, the BP2 model resulted in the closest shrinkage estimates, while the CEB-FIP model gave the least accurate predictions. Similar to the case with the creep data, the limited shrinkage prediction accuracy of the models is due to their empirical nature and to the fact that they were developed for NSC.
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Creep coefficient .
3
f' c = 10,150 psi (70 Mpa)
2.5 t o = 1 day
/f' co = 0.6 2
CEB
1.5
1
BP2
ACI
AASHTO
Curve fit
0.5
0 0.01
0.1
1
10
100 1000 10000
Time under load, days (log scale)
Figure 3.14 Creep coefficient, experimental vs predicted results
800
700
f' c = 10,150 psi (70 Mpa)
t o = 1 day
600
500
AASHTO
ACI Curve fit
400
CEB
300
BP2
200
100
0 0.01
0.1
1
10
100 1000 10000
Drying period, days (log scale)
Figure 3.15 Shrinkage strain, experimental vs predicted results
Shrinkage strain, 1 -06 .
3-40
Proposed Creep and Shrinkage Models
Based on the aforementioned comparisons, it is clear that new models are needed in order to predict creep and shrinkage deformations of HPC for early-age (less than 24 hours) loading and drying. Such models must account for the maturity of the concrete at the time of initial loading or drying, and for the effect of moist-curing on creep. The former is integral in predicting the behavior of HPC since HPC is prone to develop high initial curing temperatures, and the latter is essential in correlating laboratory test data (obtained according to ASTM specifications) to the actual behavior of HPC structural members.
Using the experimental data obtained in this research, a preliminary creep and shrinkage model was developed for HPC with concrete compressive strength between 9000 psi and 17,000 psi (62 MPa and 117 MPa). The proposed models are based on the AASHTO LRFD model, and they may be considered as an extension designed to reflect the characteristics of HPC. For creep, the proposed modifications include a stress/strength ratio factor as well as new equations for the effects of the concrete strength, the maturity at loading, and the moist curing period. For shrinkage, a maturity factor was included and both the ultimate shrinkage strain as well as the time function were modified as presented below. As more data on HPC become available, the proposed model should be further calibrated.
Creep Model
For HPC with 9000 psi fc' 17,000 psi the creep coefficient (t,to) can be calculated
as follows:
(t, to )
=
kvs k fc k H kt k km
d
(t - to )0.6 + (t - to )0.6
(3-6)
where
=
2.73
kvs = volume-to-surface ratio factor from AASHTO
t
=
26eC1(v / s) t
+
t
1.80
+ 1.77eC2(v / s) 2.587
45 + t
(3-7)
3-41
v/s =
C1 =
C2 =
k fc
=
=
fc' =
C3 = C4 = kH =
=
H =
kt
=
= k =
= = = km = = m =
d =
volume-to-surface ratio, mm (in.) 0.0142 mm-1 (0.36 in.-1) -0.02 mm-1 (-0.54 in.-1) concrete strength factor
C3 C4 + fc'
28-day compressive strength, MPa (ksi)
33 MPa (4.8 ksi)
11.32 MPa (1.645 ksi)
relative humidity factor from AASHTO
1.58 - H 120
relative humidity (%)
factor for maturity at loading
0.65
e
to
0.7 +0.57
stress/strength ratio factor from CEB-FIP
e1.5(-0.4) 1.0
for 0.4 < 0.6 for 0.4
applied stress/strength ratio at loading factor for moist curing period 1+ 0.65(1- e-0.59m )5.73
moist curing period, days
to 0.356 + 0.09to
3-42
(3-8)
(3-9) (3-10) (3-11) (3-12) (3-13)
t and to = maturity of concrete (days) at the time considered and at initial loading, respectively, calculated using the following CEB-FIP equation:
t or t = t e n
13.65-
4000 273+(T (ti
))
/
To
o
i
i=1
(3-14)
where T(ti) is the temperature (C) during period ti (days), and To is 1 C.
All the data needed for the model should be known to the engineer except for the initial curing temperature history. An accurate prediction of the initial curing temperature profile at a certain location of an HPC member is difficult because the curing temperature is influenced by many factors including mix parameters, type of cement, ambient temperature, ambient wind speed, etc. Therefore, the best way to determine the initial curing temperature profile of an HPC member is to measure it, and this data should be furnished to the engineer by the concrete producer. However, based on the curing temperature measurements obtained in this research and using Equation 3-14, the maturity (to) of G2 and G4 HPC members at the end of a 24-hour curing period (ti = 1 day) is approximately three days and four days, respectively. If the loading age is less than 24 hours, linear interpolation may be used to calculate the maturity at loading. When the loading age is more than 24 hours, the 24-hour maturity may be added to the remaining number of days up to the loading age to determine the maturity at loading.
Shrinkage Model
For HPC with 9000 psi fc' 17,000 psi shrinkage induced strains sh(t,to) can be
calculated as follows:
sh (t, to )
=
( sh )u k 'vs k ' H kto
f
t +
- (t
to -
t
o
)
0.5
(3-15)
where
(sh )u = = =
ultimate shrinkage strain 405 x 10-6 mm/mm (in./in.) for member-cured concrete 380 x 10-6 mm/mm (in./in.) for moist-cured concrete
k 'vs = volume-to-surface ratio factor from AASHTO LRFD (1998)
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=
C5 = k'H =
= =
k to
=
=
f
=
H =
to
=
t
26eC1(v / s) t
+
t
1064
-
((v / s) 923
C5 )
45 + t
3.7 mm-1 (94 in.-1)
relative humidity factor
140 - H 70
3(100 - H ) 70
for H < 80% for H 80%
factor for maturity at the beginning of drying
0.67
e
to
4.2 +9.45
23 days relative humidity (%) maturity at the start of drying
(3-16) (3-17)
The predictions of the proposed model were compared to those obtained by the four other models. The data used in the comparison included the 19 data sets obtained in this research and an additional 17 data sets reported in the literature (Huo, 1997; and Mokhtarzadeh and French, 2000). The results of the comparison are plotted in Figure 3.16 in terms of the overall coefficient of variation which is calculated for all data sets combined. The results of the comparison for each individual data set can be found in Shams and Kahn (2000) and Saber(1998). With respect to the creep data, Figure 3.16 indicates that the proposed model is superior to all the other models with a coefficient of variation of only 18.2%. The results of the comparisons for the individual data sets indicated that the proposed model is much more appropriate for HPC loaded at ages less than 24 hours. In addition to its improved accuracy, simplicity was maintained in the proposed model by only requiring information that the engineer should know beforehand.
3-44
The proposed shrinkage model also appeared to render better predictions for HPC. However, the statistical analysis results in Figure 3.16 indicate that the proposed model improved the predictions by only 7% compared to the BP2 model.
, %
200 180 160 140 120 100
80 60 40 20
0 ACI AASHTO CEB
Creep Shrinkage (sh )u
BP2 Proposed Model
Figure 3.16 Proposed creep and shrinkage models compared to other models.
Conclusions Regarding Creep and Shrinkage
It was concluded that the creep coefficient of HPC is inversely proportional to the concrete compressive strength and to the age (maturity) at loading. The creep coefficient also is independent of the applied stress/strength ratio up to a stress level of 0.4, and is directly proportional to the applied stress/strength ratio for stress levels between 0.4 and 0.6. The creep models of ACI Committee 209, CEP-FIP code, Bazant and Panula, and AASHTO LRFD overestimated creep of HPC with the AASHTO code model giving the closest predictions.
As compared to normal strength concrete drying under the same conditions, HPC exhibits 31% to 51% less shrinkage strains, and one-half of that shrinkage takes place during the first two weeks of drying. Shrinkage of HPC was overestimated by ACI Committee 209 and AASHTO models, and it was underestimated by the CEB-FIP model and the simplified Bazant and Panula's model with the latter giving the closest predictions.
3-45
For HPC with 9000 psi fc' 17,000 psi , the proposed models are recommended for predicting creep and shrinkage. These models were shown to predict the creep and shrinkage results of this research and of other studies with overall coefficients of variation of 18.2% and 37%, respectively. However, as more data on HPC become available, the proposed models should be further calibrated.
3-46
Chapter 4. Type II Girders: Transfer and Development Length of 0.6-in. Strand
4.1 Introduction
The purpose of this portion of the experimental research program was to verify that the
transfer and development length of 0.6-inch (15-mm) diameter prestressing strand were less than
calculated by the current American Association of State Highway and Transportation Officials Specifications (AASHTO 17th Edition, 1996) when used in high performance concrete (HPC)
bridge girders. The scope was limited to testing four AASHTO Type II girders, two made from
10,150 psi (70 MPa) and two from 14,500 psi (100 MPa) design strength concretes. Transfer length (lt) is defined according to the American Concrete Institute (ACI 318-99)
as "the distance over which the strand must be bonded to the concrete to develop the effective prestress, fse, in the strand." The effective prestress ( fse ) is the stress in the reinforcement after
allowance for all prestress losses. The flexural bond length is defined as "the additional length over which the strand must be bonded so that a stress fps may develop in the strand at nominal strength," where fps is the stress in the prestressed reinforcement at nominal strength. Development length (ld) is defined according to ACI as the sum of the transfer length and the
flexural bond length. Figure 4.1 presents an idealized strand stress profile in a pretensioned
element under applied load.
The ACI 318-99 code suggests the transfer length can be calculated from the following expression: lt = fseD / 3 . D (in.) is the diameter of the prestressed reinforcement. The American
Association of State Highway and Transportation Officials specifications (AASHTO 17th Edition, 1996) suggest the transfer length be taken as 50 times the strand diameter (50*D). Both
the ACI and AASHTO codes suggest that the development length can be calculated from the
following expression:
ld
=
fps -
2 3
fse D
where fps (psi), fse (psi), and D (in.) were defined previously.
(4-1)
4-1
Steel Stress
Transfer fps Length
fse
Development Length Flexural Bond Length
Increase in strand stress due to applied loads.
0
lt
Distance from free end of strand
ld
Figure 4.1 Idealized strand stress profile in a pretensioned element under applied load. (Graph built using properties determined in this study.)
4.2 Type II Girder Design, Construction, and Instrumentation Four AASHTO Type II girders were cast using two different grades of HPC, as defined
by Goodspeed, et al. (1996). Two girders were cast with a Grade 2 HPC, 10,150 psi design strength; and two were cast using a Grade 4 HPC, 14,500 psi design strength. Each girder was 33 ft 5.5 in. long. Composite concrete decks were cast atop all girders using a Grade 1 HPC, 7,250 psi design strength. The decks were 30 in. wide and 8 in. deep. Figures 4.2 and 4.3 illustrate the girders.
Ten 0.6-inch diameter, low-relaxation prestressing strands, with a minimum strength of 270 ksi, were used to pretension each girder. All of the strands were straight, fully stressed and bonded along the entire girder length. The strands were stressed to an initial stress, fpt, of 190.4 ksi. Eight strands were placed in the bottom flange, centered 3 in. above the bottom of the section, with a center-to-center spacing of 2 in. Two strands were placed in the top flange, centered 2.5 in. below the top of the girder section, spaced 2 in. apart.
The concrete decks were cast to model bridge girders and to maximize the strain levels in the strands during the development length tests. Longitudinal shrinkage and temperature reinforcement in the deck consisted of two #4 bars placed 7.5 in. from each side of the composite deck.
4-2
762
190
190
203
1041
Grade 1 HPC Deck
Shrinkage Reinforcement #13 Bar
Shear Reinforcement #13 Stirrups
AASHTO Type II Girder
Eight 15 mm diam. strands 76
Figure 4.2 Typical cross section of Type II girders. Dimensions in mm.
4 @ 100 50
17 @ 150
10.2 m (33'-5 1/2")
30 @ 225
4 @ 100 50
South End
50
23 @ 100
Elevations of Girders G2A and G4A
10.2 m (33'-5 1/2")
48 @ 150
North End
6 @ 100 50
1 inch = 25.4 mm
Elevations of Girders G2B and G4B
Figure 4.3 Elevation of Type II girders showing stirrup layout. Dimensions in mm.
4-3
The transverse shear reinforcement consisted of C-shaped Grade 60, #4 (M13) deformed bars with a yield strength of 75 ksi. Figure 4.3 illustrates the stirrup spacing. Each girder was labeled according to the grade of girder concrete: G2A, G2B, G4A, and G4B; each end was labeled either North (N) or South (S).
During construction, prestressing force in each strand was measured using a load cell; the average stress prior to transfer (fpt) was 190.4 ksi. Transfer length was determined using two techniques, concrete surface strain (CSS) at the level of the bottom strands and strand end slip.
The CSS was measured using DEMEC gages with an 8 in. gauge length. Embedments spaced 2 in. on-center were cast in each girder on each side of the bottom flange for a distance of 12 ft from each end. Strand end slip was measured at the prestressing plant using dial gages that were attached to the bottom strands prior to transfer of the prestress. Spring loaded rod potentiometers (SLRP's) were attached to the bottom strands to measure end slip during each of the development length tests.
To measure long-term deformations, DEMEC gauge points also were embedded in each girder at midspan at the level of the top strands, girder centroid, and bottom strands. Further, a vibrating wire strain gage was located at the level of the bottom strands.
DEMEC gage points were epoxy bonded to the exterior of the girder and the composite deck to measure the strain profile beneath the load point for each development length test. During the development length tests, the girder deflection was measured at four locations using linear potentiometers.
The four Type II girders were cast at the Standard Concrete Products precast/prestressing plant, Atlanta, Georgia. Normal procedures were used for construction. The forms were removed 20 hours after casting, and initial measurements were taken. The strands were flame cut 22 hours after casting when the concrete had reached nearly 100 percent of its design strength as measured using match-cured (accelerated cured) cylinders. The composite decks were cast atop each girder about two months after the girders were fabricated.
4-4
4.3 Material Properties
Table 4.1 presents the mix proportions for the three grades of HPC, and Table 4.2 presents the concrete material properties. The 4x8-in. test cylinders were cured in an insulated box, as discussed in Chapter 3 to match the internal temperature of the girders. Four segments of 0.6-inch diameter strand were tested and found to possess the following average characteristics: Eps = 28,625 ksi, fpy = 257.5 ksi, and fsu = 280.0 ksi at an su = 0.053 (in./in.). The transverse shear reinforcement was #4 (M13) Grade 60 (60 ksi) deformed bar having a fy of 75 ksi, although 60 ksi was used in all shear calculations per ACI 318-99.
Table 4.1 Mix Designs for HPC Type II Girders
Constituent Ingredients Design Strength (ksi): Cementitious (lb/yd3):
Type I Flyash, Class F Silica Fume
Aggregate (lb/yd3): Fine, Natural Sand Coarse, #67 Granite Coarse/Fine Aggregate Ratio
Net Water (lb/yd3): Water/Cement Ratio Water/Cementitios Materials Ratio Chemical Admixtures (oz/yd3): Set Retarder Superplasicizer* Air Entrainment (Darex)
Grade 1 Grade 2 Grade 4 7,250 10,150 14,500
650
650
835
100
100
100
15
35
80
1150 1700 1.48 240 0.37
0.31
1000 1750 1.75 215 0.33
0.27
900 1800 2.00 235 0.28
0.23
20
20
27
91
200
250
4
20
0
4-5
Table 4.2 Summary of Test Data
Property design fc' girder fc' girder 56 day (ASTM) fc' girder 56 day (INS) fr girder 56 day (ASTM) fc' girder at testing (ASTM) fc' girder at testing (INS)
design fc' deck fc' deck 56 day (ASTM) fc' deck at testing (ASTM)
units G2AN G2AS* G2BN G2BS
(psi) 10,000
10,000
(psi) 15,465 17,972 15,465
(psi) 13,430 16,106 13,430
(psi)
1,003 1,348 1,003
(psi) 16,251 18,947 16,251
(psi) 15,169 16,768 15,169
(psi) 7,000
7,000
(psi) 7,580
6,947
(psi) 8,271
8,364
Main Span
(in) 389.75
270 321.25 390.25
Shear Span
(in)
114
74
82
69
Embedment Length
(in)
120
80
88
75
Test Failure Mode
Flex. Flex. Flex. S/Bond
Avg. End Slip during
(in)
0.010 0.033 0.037 0.396
testing Experimental Mcr Predicted Mcr (exp. fr) Exp. Mcr/Pred. Mcr (exp. fr) Predicted Mcr (code fr) Exp. Mcr/Pred. Mcr (code. fr) c max (exp. top fiber strain)
(k*in) (k*in) (k*in)
(in/in)
13,655 14,460 0.944 13,780 0.991
14,773 16,220 0.911 14,190 1.041
15,137 14,470 1.046 13,790 1.098
13,480 14,470 0.932 13,790 0.978
0.0029 0.0229 0.0031 0.0023
Status of Deck Concrete
Crush Crush Crush Crush
ps max (exp. strand strain) fps max (exp. strand stress ) Experimental Mmax Predicted Mn Mmax / Mn Experimental Vcw Pred. Vcw (mid shear span) Exp. Vcw / Pred. Vcw Experimental Vmax ACI/AASHTO (16th) Vn
(in/in) (ksi) (k*in) (k*in)
(kip) (kip)
(kip) (kip)
0.031 271.7 20,696 19,990 1.035
177 138 1.28 182 232
0.037 273.8 21,128 19,990 1.057
210 138 1.52 286 279
0.038 274.3 21,488 19,870 1.081
191 138 1.39 262 279
0.026 270.0 20,222 19,870 1.018
211 138 1.53 293 307
Vmax / Vn ACI AASHTO (LRFD) Vn Vmax / Vn AASHTO (LRFD)
0.78 1.03** 0.94 0.95
(kip)
182
235
256
283
1.00** 1.22** 1.02** 1.04
* Misplaced G4 girder concrete.
Flex. = Flexural Failure
S./Bond = Shear/Bond Failure
Eight bottom strands ruptured.
Shear failure occurred at less load than calculated capacity. (unconservative)
** Incorrectly predicted shear failure.
G4AN 14,000 17,972 16,106 1,348 18,947
16,768 7,000 7,529 8,459 307.75
101 105 Flex. 0.009
13,917 16,040 0.868 14,040 0.991
0.0027 Crush 0.053 279.9 21,693 19,970
1.086 179 138 1.30 215 232 0.93 169
1.27**
G4AS
391.75 74 80
S/Bond 0.367
11,906 16,040 0.742 14,040 0.848
0.0021 Intact
0.020 267.6 18,772 19,970 0.940
177 138 1.28 254 279 0.91 239 1.06
G4BN 14,000 17,972 16,106 1,348 18,947
16,768 7,000 8,910 10,632 327.5
89 93 Flex. 0.055
13,489 16,450 0.820 14,390 0.937
0.0032 Crush
0.051 279.2 21,558 20,300 1.062
179 138 1.29 242 279 0.87 264 0.92
G4BS
393.5 61 65
S/Bond 0.273
13,628 16,450
0.828 14,390
0.947
0.0012 Intact
0.022 268.4 20,045 20,300 0.987
251 138 1.82 329 307 1.07 253 1.30
4-6
4.4 Transfer Length
The strand end slip method for measuring transfer length proved unreliable because (1) over half the gages broke during flame cutting and (2) the remaining gages gave widely varying results. The concrete surface strain (CSS) technique using the DEMEC gages provided useful data that was much more reliable and consistent. Transfer length was calculated using the 95 percent average maximum strain (95% AMS) method discussed by Russell and Burns (1992). Figure 4.4 gives the smoothed concrete surface strain data for specimen G2B North while Figure 4.5 illustrates the DEMEC measurement. The plot is bilinear with the first region being the transfer region and the second region being the constant strain plateau that extends the entire length of the girder until it reaches the transfer region at the other end. The surface strains for each girder resemble those shown in Figure 4.4. The transfer length was found to increase up to 7 days and then level off. Table 4.3 presents the transfer length results at release and at 7 days. The transfer length averaged 17.6 in. and 14.6 in. for the Grade 2 and Grade 4 HPC girders, respectively.
Transfer Length Determined from 95% AMS Method Girder G2B-North
1400 Transfer Length = 457
1200
Strain ()
1000 800 600 400
Defined Constant Strain Plateau
Averaged, Smoothed Strain Profile
95% AMS
200
0
0
500 1000 1500 2000 2500 3000 3500
Distance from Girder End (mm)
Smoothed
95% AMS
Figure 4.4 Transfer length using 95% AMS method for specimen G2B North (mm)
4-7
Figure 4.5 DEMEC transfer length strain measurements immediately after cutdown
Table 4.3 Transfer Length of 0.6-in. strand in HPC Girders (in.)
Time After
Prestress
G2A
Transfer (days) North South*
0.1
16.0 17.0
6.7
17.0 15.0
* Misplaced G4 girder concrete.
1 inch = 25.4 mm
G2B North South 16.0 16.0 17.5 18.5
G4A North South 13.0 14.5 14.5 14.0
G4B North South 15.0 12.5 13.5 11.5
In addition, 12 direct pull-out tests were performed on unstressed samples of strand that were embedded 18 in. in concrete blocks made from the G2 and G4 concretes as illustrated in Figure 4.6. The tests indicated that the yield stress of the strands could be achieved, and that the strands exhibited good bond (Logan, 1997 and Zia and Mostafa, 1997).
4-8
Figure 4.6 Strand pull-out test
4.5 Girder Test Results To determine the minimum strand development length, each girder was simply
supported, and a point load was applied at an embedment length from one end, as illustrated in Figures 4.7 and 4.8. The embedment length was varied when each end of each girder was tested, as listed in Table 4.2. The minimum embedment length at which the girder failed in a flexural mode while reaching its theoretical flexural capacity with less than 0.1-inch bond slip was defined as the development length. A complete discussion of the development length tests can be found in work done by Dill (2000).
Each girder was loaded monotonically in small deflection increments of about 0.1 inch. The tests were stopped when the girder failed in either a shear/bond (S/bond) or flexural mode. Shear/bond failures were evident when large inclined cracks were present, when the end slip of the bottom strands was greater than 0.25 in., and when slip continued to increase with no increase in load carrying capacity. Flexural failures consisted of ductile behavior and yielding of the prestressed reinforcement. The tests were stopped when the deck concrete crushed at a strain of about 0.003.
4-9
Embedment Length
Loading Apparatus
Bearing Plate (Point Load)
Shear Span
HPC Composite Girder
Span Length
Deflection Measuring Apparatus
Figure 4.7 Development length test set up
Figure 4.8 Development length test of G2B-South. Note shear crack becoming a bond failure crack at the level of the prestressing strands.
Following is a discussion of the development length tests in regard to flexural behavior, shear behavior, and bond behavior and development length status. Table 4.2 presents the results of the development length tests including span, shear span, embedment length, failure load (Pmax), and failure mode.
4-10
4.5.1 Flexural Behavior The flexural behavior of the girders was examined by comparing theoretical and
experimental moment-curvature relationships. A predicted moment-curvature relationship was based on Lin and Burns (1981) and assumed elastic behavior to cracking and a parabolic concrete stress-strain response with a maximum compressive strain of 0.003. Figure 4.9 illustrates the predicted and experimental moment-curvature relationships for test G4AN. For this test the maximum concrete strain was measured as 0.0035. Shortly after the concrete crushed, the bottom strands ruptured. Other moment-curvature relationships were similar to Figure 4.9. The experimental curvature and strain profile were obtained from the external strain readings at various elevations at the section under the point load.
Moment (kN*m)
2500 2000 1500 1000 500
0 0
0.01 0.02 0.03 0.04 Curvature (rad/m)
Predicted
Experimental
Figure 4.9 Moment curvature predicted and experimental results for G4A-North
The nominal moment capacity was calculated based on strain compatibility with a top fiber strain of 0.003. The flexural failures in this study exhibited an average value for cu of 0.00296. Table 4.2 presents a summary of the maximum strains in the deck and the prestressing reinforcement. (The top fiber strains were extrapolated from the strain profile data.)
4-11
The cracking moments (Mcr) were calculated using the experimental modulus of rupture and using the traditional value, fr = 7.5 f ' c , where fr and fc' are in psi. This was the only time f ' c was allowed to exceed 100 psi (ACI 318-99). Calculations based on the experimental modulus of rupture overestimated the actual cracking moment for 7 of the 8 tests and overestimated the cracking moment by as much as 26 percent. A calculation based on the code-specified modulus of rupture was very accurate for 7 of the 8 tests, but overestimated one test (G4AS) by 15 percent. This comparison suggests that the existing code calculation of modulus of rupture may be more accurate than the modulus of rupture determined from standard modulus of rupture tests for HPC. Table 4.2 presents a summary of the cracking moments found in testing the girders. The code specified modulus of rupture, fr = 7.5 f ' c , appears to be accurate for the girders in this study. In addition, the allowable tension limit of 6 f 'c appears to be appropriate even when the f ' c is allowed to exceed 100 psi.
All of the tests strained the prestressing reinforcement to at least twice the yield strain. Although some girders failed in a shear/bond mode, the failure loads were near those calculated based on nominal moment capacity of those girders. The average experimental stress in the bottom strands (273.1 ksi) was very close to the strain compatibility calculated stress of 275 ksi. G4A North was the only test that strained the bottom reinforcement to the point of rupture.
The calculations of nominal moment (Mn) were very close to the experimentally observed values. The calculated moment capacity was 3 to 9 percent less than the experimental ultimate moments. Only G4A South and G4B South did not reach the calculated moment capacity because both tests failed in a shear/bond mode. Table 4.2 presents a summary of the maximum observed moments and the calculated moment capacities. As an example, Figure 4.10 shows the flexural failure of test G4A North.
4.5.2 Shear Behavior Shear/bond failures occurred in three tests after substantial end slip of the bottom strands.
In general the strands would begin to slip, which would decrease the precompression of the girder concrete and permit widening of the diagonal shear cracks. This widening would lead to less concrete shear resistance and failure of the girder.
4-12
Figure 4.10 Flexural failure of Type II HPC girder G4A-North
In this study, cracking shear (Vcw) was calculated according to the ACI 318-99 and AASHTO (16th Edition, 1996) codes. (Vcw was always less than Vci.) The f ' c was limited to 100 psi. Table 4.2 presents the calculated cracking shear, at mid-shear span, and the experimentally observed cracking shear. The calculated value was found to be from 28 to 82 percent less than the observed shear force. Even when the f ' c was allowed to exceed 100 psi, the calculated cracking shears were conservative by at least 18 percent.
Nominal shear resistance (Vn) was calculated according to both the ACI 318-99 and AASHTO (16th Edition, 1996) method and the AASHTO (LRFD 1998) method. Both techniques produced similar results, but the AASHTO (LRFD 1998) method appears to be more conservative and more accurate. Table 4.2 presents the calculated nominal shear
4-13
capacities and the experimental maximum shears. The f ' c was limited to 100 psi, and fy was limited to 60 ksi.
The ACI 318-99 and AASHTO (16th Edition,1996) method correctly calculated 1 of the 3 shear failures, but it gave a 5 and 9 percent greater capacity for the other two tests that exhibited shear/bond failure. In addition, this method incorrectly calculated shear failure for test G2AS. The AASHTO (LRFD 1998) method correctly calculated the three shear failures, but it calculated 27 percent less capacity for test G4AN in which little shear cracking was observed and in which the bottom strands were ruptured. As an example, Figure 4.11 shows the shear/bond failure of test G4B South. The ACI 318-99 and ASSHTO (16th Ed.,1996) method appear appropriate for girders in which end slip is restricted.
Figure 4.11 Shear failure of Type II HPC girder G4B-South. Note horizontal cracking at the level of the prestressing strands.
4-14
4.5.3 Bond Behavior and Development Length The concrete surface strain was measured at the level of the bottom prestressing strands
during each of the development length tests. The measured strain was added to the effective prestressing strain to determine the strain in the prestressing reinforcement. Figure 4.12 presents the strain in the bottom strands for test G2A North, which was typical of the flexural failures. A strain diagram that represented a shear failure gave higher strains near the end of the girder where the large shear cracks crossed the bottom strands.
0.04
Average Strain (mm/mm)
0.03
Load 1,067 kN
0.02 0.01
Load 1,027 kN Load 952 kN
Load Point
0 0 0.5 1 1.5 2 2.5 3 3.5
Distance From End (m)
Figure 4.12 Strain in the bottom strands in G2A-North. (Curves represent strains at prestress forces corresponding to strains of 0.01, 0.02, and maximum strain) 1 kip = 4.448 kN 39.37 in = 1 m
End slip of the bottom strands was monitored during the development length tests and was a very good indicator of the failure mode. The three tests that experienced shear/bond failure modes had average maximum end slips of 0.40, 0.37, and 0.27 in. In general, if extensive shear cracking occurred in the transfer region, the strands experienced significant end slip and a development length failure resulted. Similarly, for the tests that exhibited mild shear cracking, with small or no cracking in the transfer region (within 16 in. of the support), the failure mode was flexure. The five tests that exhibited flexural failure had an average end slip
4-15
of less than 0.06 in. It was noted that the outer strands slipped 2 to 10 times more than the inner strands.
Figure 4.13 is a graph of end slip versus embedment length for the eight tests. As seen in the graph, 2032 mm (80 inches) was the critical embedment length that separated the shear/bond failures from the flexural failures; therefore, the development length was determined experimentally to be 2032 mm (80 inches). Both the Grade 2 and Grade 4 concretes in this study had the same bond characteristics.
Avg. End Slip (mm)
12 Shear/Bond Failures (3)
9
6 Flexural Failures (5)
3
0
1.5
2
2.5
3
Embedment Length (m)
Average G2 G4
Figure 4.13 Average end slip vs. embedment length. 1 in = 25.4 mm
Two tests were performed with an embedment length of 80 in. (2032 mm) where the sections had the same shear reinforcement: G2AS and G4AS. Test G2AS was a ductile flexural failure with minimal end slip, whereas test G4AS exhibited a shear/bond failure. Test G4AS failed at an applied shear about 13 percent less than G2AS. The onset of bond failure in G4AS decreased the shear resistance and allowed extensive shear cracking to occur. Consequently the girder prematurely failed. The two different failure modes at this 80 in. (2032 mm) embedment length helped establish this length as the minimum development length for the 0.6-inch (15-mm) diameter prestessing stand.
4-16
4.6 Comparison with Other Research Past research on transfer length of prestressing strand has led to the development of
several proposed transfer length expressions that account for various factors such as the level of prestress (jacking, initial, and effective), strand diameter, concrete compressive strength (initial and 28-day), and the concrete elastic modulus. Table 4.4 presents several expressions to calculate transfer length. Table 4.4 also includes ratios of calculated strand transfer length to the experimental strand transfer lengths. The equation presented by Mitchell et al. (1993) produced the best prediction of strand transfer length.
Similarly, past research has led to several proposed development length expressions. Concrete strength, strand diameter, stress in the steel at nominal strength, and effective prestressing stress are the major factors used in the proposed development length expressions. Table 4.5 presents expressions to calculate development length. Table 4.5 also includes ratios of calculated strand development length to the experimental development length. The equations by Zia and Mostafa (1997), Lane (1998), and the ACI 318-99 and AASHTO code equations conservatively calculated the development lengths within 21 percent of that found experimentally.
4.7 Conclusions Regarding Transfer and Development Length The 0.6-in. diameter prestressing strands showed good bond and development
characteristics in high performance concrete with compressive strengths less than 14,500 psi and, therefore, are recommended for use in pretensioned HPC bridge girders. Direct pull-out tests indicated reliable bond. The transfer length was determined to be 41 percent less than that calculated by the current AASHTO (16th Edition, 1996) specification of 50 times the strand diameter for the Grade 2 concrete, and it was 51 percent less than calculated for the Grade 4 concrete. For both grades of concrete, the development length was determined to be 20 percent less than that calculated by the current AASHTO Standard (1996) code equation. The current AASHTO Standard (1996) provisions for transfer and development length are recommended for HPC girders with concrete strengths less than 14,500 psi because of their simplicity and accuracy.
4-17
Table 4.4 Comparison of Predicted and Experimental Transfer Lengths
Source
AASHTO (16th) AASHTO (LRFD) ACI 318-99
Buckner (1995)
Transfer length expression
lt = 50D lt = 60D
lt = fseD
3
lt
=
1250 fsiD Eci
HPC Age (day)
Grade 2
(in.) 30 36
Calculated Experimental
1.70 2.04
Grade 4
(in.) 30 36
Calculated Experimental
2.05 2.46
34.1
1.94
34.1
2.33
1 23
1.30
21.8
1.49
Lane
(95% Confidence) f'ci limited to
l t
=
4
fptD f 'c
-5
28 40.7
2.31
40.7
2.79
10 ksi
Martin and Scott
lt = 80D
48
2.73
48
3.29
Mitchell et al.
lt = 0.33 fsiD
3 f ' ci
1
17.6
0.99
16.1
1.10
Russell and Burns
lt = fseD 2
51.6
2.93
51.6
3.53
Zia and Mostafa f'ci limited to
l t
=
1.5
fsiD f ' ci
-
4.6
16
0.91
16.1
1.10
8 ksi
Zia and Mostafa
l t
=
1.5
fsiD f ' ci
-
4.6
1
8.5
0.48
6.5
0.44
Georgia Tech lt = 17.6 in. and 14.6 in., for Grade 2 and 4 HPC respectively
4-18
Table 4.5 Comparison of Predicted and Experimental Development Length
Source
AASHTO (16th) and ACI 318-99 Buckner (1994)
Deatherage et al.
Lane (95% Confidence) fc' limited to 10 ksi Mitchell et al.
Development length expression
ld = ( fps - 2 fse)D 3
ld = fsiD + ( fsu - fse)D
3
1.0 [ = (0.6 + 40ps)] 2.0
ld = fsiD + 1.5( fsu - fse)D
3
l d
=
(4
fptD f 'c
-
5)
+
( 6.4(
fsu f
- 'c
fse)D
+ 15)
Grade 2
(in.) 96.2
166.2
133.8
97.1
ld = 0.33 fsiD
3 f ' ci
+
( f su
-
f se )D
4.5 f 'c
55.4
Calculated Experimental
1.20 2.08
1.67 1.21
0.69
Grade 4
(in.) 96.1 166.1
133.7 97.1
50.8
Calculated Experimental
1.20 2.08
1.67 1.21
0.63
Zia and Mostafa f'ci limited to
55.2 MPa
ld
=
(1.5
fsi f ' ci
D
-
4.6)
+ 1.25(
fsu
-
fse)D
97.0
1.21
97.0
1.21
Zia and Mostafa
ld
=
(1.5
fsi f ' ci
D
-
4.6)
+ 1.25(
fsu
-
fse)D
89.7
1.12
87.7
1.10
4-19
Chapter 5. HPC Demonstration Bridge Evaluation
5.1 Introduction A high performance concrete highway bridge was constructed in Georgia as part of the
research program. The precast, pretensioned girders had a specified strength of 10,150 psi (70 MPa) while the 8-in. thick cast-in-place bridge deck had a specified strength of 7,250 psi (50 MPa). The skewed bridge crossed Interstate 75 in Henry County Georgia, approximately 15 miles southeast of Atlanta, GA. The total length was approximately 353 ft (107.5 m) long by 47 ft (14.4 m) wide and consisted of four simply supported spans as illustrated in Figure 5.1. The longer of the two exterior spans was 50 ft-1 in. (15.26 m) center-of-bearing to center-ofbearing and used Type II AASTHO girders spaced 7 ft-3 in. (2223 mm) on center. The two interior spans were 124 ft-1 in. (37.82 m) center-to-center and used Type IV AASHTO girders at the same spacing. Comparative design using 6,000 psi (41.4 MPa) concrete would have required an extra five girders per span at a 5 ft-3 in. (1,600 mm) spacing. The HPC bridge was constructed at an approximate cost of $51.38 per square foot of finished bridge deck, compared with the average cost of $56.65 for a typical, normal strength concrete precast, prestressed composite bridge in Georgia. Two girders in each of the first two spans of this bridge were instrumented to monitor their internal and concrete surface strains, deflection, and loss of prestress.
Grades 1 and 2 High Performance Concrete as defined by Goodspeed et al. were used as a basis for developing the specifications for the bridge deck and girders, respectively. The original grade 1 bridge deck mix had a specified compressive strength of 7,250 psi at 56 days and a maximum rapid chloride permeability of 2,000 coulombs; the grade 2 bridge girder mix had a specified compressive strength of 10,150 psi and a maximum permeability of 3,000 coulombs.
Figures 5.2 and 5.3 illustrate the completed bridge structure. Detailed descriptions of the bridge performance are provided by Slapkus and Kahn (2002) and by Lopez and Kahn (2005).
5.2 Girder Construction and Instrumentation AASHTO Type II and IV HPC Girders for Phase 1 were constructed in the summer of
2001 at Standard Concrete Products plant in Atlanta, Georgia. The concrete used to cast these
5-1
girders was specified as a "Class AAA HPC" concrete by the Georgia DOT. The mixes are given in Table 5.1.
SPAN 1
53'-4"
TOTAL LENGTH OF BRIDGE = 353'-0"
SPAN 2 127'-2"
127'-2"
45'-1"
NORTH
58.75
59.5
BENT 1
BENT 2
47'-0"
61
47'-0"
TO LOVEJOY
62.5
63
CONSTRUCTION
JOINT IN DECK BENT 3
BENT 4
BENT 5
TO McDONOUGH
PHASE 1 PHASE 2
Figure 5.1 Bridge plan, Phases 1 and 2
Span 4, Type Span 3, Type
II girders
IV girders
Span 2, Type IV girders
Span 1, Type II girders
Figure 5.2 Completed bridge structure. 5-2
Diaphragm Girder 2.1
Bent 3
Girder 2.4
Girder 2.5
Figure 5.3 View looking east showing instrumented girders 2.4 and 2.5 of Phase 1
Table 5.1 Girder concrete mixes
Average components, Average components, Approved
Component Type
girders 1.4 and 1.5 girders 2.4 and 2.5 mix design
Cementitious material (lb/yd3):
cement, type I
809
805
796
Flyash, type F
102
103
98
Silica Fume, Force 10,000
75
75
70
Aggregates (lb/yd3):
fines, Brown Brothers #2 sand
985
978
965
coarse, Vulcan, #67 stone
1820
1820
1837
Water (lb/yd3):
261
260
237
Chemical Admixtures (oz/yd3):
AEA, Daravair 1000
10
9
7
Water reducer,WRDA 35
35
35
35
HRWR, Adva 100
145
130
169
Retarder
0
24
0
5-3
The 270 ksi low relaxation prestressing strand had a 0.6-inch diameter, an average modulus of elasticity of 29,600 ksi, an average yield stress of 259 ksi, and average ultimate stress of 279 ksi, and an average ultimate strain of 4.7 percent.
The Type II and Type IV girder designs are illustrated in Figures 5.4 and 5.5, respectively.
Length = 51'-2.5" (15,609mm)
10" DETAILS SYMMETRICAL ABOUT MIDPOINT
7 SPACES AT 20"
16 SPACES 4 SPACES #5 Stirrup AT 8.5" AT 4" Details
2 STRANDS Diaphragm mounting holes
2 STRANDS
6 bottom flange "doghouse" hoops at 12"
Ctr to Ctr brg = 50'-1" (15,259 mm)
4 STRANDS 6 STRANDS
Elevation
Section @ Midspan
Section @ End
Figure 5.4 Type II girder design, used for Span 1 and Span 4.
Load cells were placed on strands at the anchorage location for both types of girders. The average force in the Type II girders was 41.92 kips and in the Type IV girders was 41.16 kips.
Figure 5.6 illustrates the 6-inch gauge length vibrating wire strain gauges (VWSG) being installed in the girders. The VWSG were installed at quarter span and midspan locations of each girder; they were positioned at the center of the top layer of strands, at the center of the bottom layer, and at the top of the bottom group of strands. Thermocouples were also installed to measure early age temperatures during curing; they were located in the center and edge of the bottom flange and in the web. Detachable mechanical (DEMEC) inserts were bolted to the sides of the girder forms so that external strain measurements could be taken after the girders were cast. Camber was measured to the nearest 0.020 inches with a scale and using a taught wire.
5-4
Length = 125'-3" (38,177 mm)
DETAILS SYMMETRICAL ABOUT MIDPOINT
10"
21 SPACES AT 20"
16 SPACES 4 SPACES Details
AT 8.5"
AT 4"
2 STRANDS
Diaphragm mounting holes
12 - DRAPED STRANDS
6 bottom flange "doghouse" hoops at 12"
Ctr to Ctr brg = 124'-1.5" (37,827 mm) 1905 10 STRANDS
4 STRANDS 8 STRANDS 6 STRANDS
Elevation
Section @ Midspan
Section @ End
Figure 5.5 Type IV girder design, used for Span 2 and Span 3.
VWSG
Thermocouple Figure 5.6 Installing VWSG in Type II girder (left) and Type IV girder (right)
5-5
Concrete placement took place for the Type II girders on August 28, 2001 and on September 6, 2001 for the Type IV girders. Girders 1.4 and 1.5 were poured in two lifts and 2.4 and 2.5 were poured in three. Each girder was filled continuously until it was completed before the pour proceeded to the next girder. Concrete was consolidated with spud vibrators. Four full batches and two partial batches were needed to pour girders 1.4 and 1.5; twenty eight batches were needed to pour girders 2.4 and 2.5.
Immediately after the pour, the girders were screeded, finished, and roughened to an amplitude of approximately 1/4 in. (6mm) by raking. After a few hours, tarps were placed over the forms where they remained until just before release. The girders were not steam cured. The prestressing strands were cut the following day.
High heats of hydration were recorded for the girders; temperatures during curing over the first 24 hours were in excess of 160oF (71oC). In the morning after the girders were poured and before release of the prestressing strands, the internal girder temperatures were approximately 40oF lower than the peak temperatures recorded during hydration. Two or three small vertical cracks were seen on each of the top of the Type II girders. However, many large, full-depth cracks (as much as 0.075 in. wide) were observed for the Type IV girders immediately after side forms were removed. Cooling of the girder was considered as the cause of the cracking; although the girder tried to contract, it was restrained by the tensioned strands. After the strands were cut, all cracks closed. Calculations of stresses according to service loads showed that the section was always in compression. Zia (1993) suggested that autogenous healing mended such cracks and that the girders' performance would not be affected. Figure 5.7 shows a diagram of the location and sizes of thermal cracks observed in the two instrumented Type IV girders. Figures 5.8 and 5.9 give the typical temperature-time curves for the control specimens and girder during curing.
For each girder, VWSG and DEMEC measurements were made immediately after the forms were removed and after the strands were cut. After the strands were cut, all thermal cracks closed and were not visible. The Type IV girders were wet cured for one week to assure healing of the thermal cracks.
As shown in Figure 5.10, cracking occurred at one end of each Type IV girder after cutdown. This cracking resulted from restraint to sliding. It affected the transfer length. As
5-6
discussed in Chapter 6, Teflon pads were used in other Type IV girders which eliminated this cracking.
0.030"
0.040"
0.005" 0.01" 0.030"0.040" 0.075"
0.060" 0.030"
Bent 2 End
Girder 2.4
8' 10' 6' 10' 10' 6' 12'
26'
12'
Bent 3 End
Crack width, top of girder, (typ)
0.013" 0.040"
0.016" 0.030"
0.013"
0.02"
0.040" 0.050" 0.050"
0.005" 0.007" 0.025" 0.013"0.009"
Bent 2 End
Girder 2.5
Bent 3 End
9'
17'
17'
13' 4' 3' 8' 8' 3' 14' 10' 9' 3' 4'
Figure 5.7 Initial thermal cracks observed on girders 2.4 and 2.5 just after removal of forms.
180
160
140
Temperature (oF)
120
100
Insulated 6x12 cylinders
Insulated 4x4x14
80
Beams
Match cure 4x8 cylinders
60
0
4
8
12
16
20
24
Time after casting (hours)
Figure 5.8 Thermal measurements of control specimens for Type II girders
5-7
180 160 140
TC 1.B cl 4
TC 1.B cl 3 TC 1.B cl 1
TC 1.B cl 2
Temperature (oF)
120
100
80
TC 1.4cl 1
TC 1.4cl 2
TC 1.4cl 3
TC 1.4cl 4
60
0
4
8
12
16
20
24
Time after casting (hours)
Figure 5.9 Thermocouple (TC) measurements at midspan of Type II girder 1.4
5-8
web cracks begin at bottom harped strand and continue into bottom flange
bottom flange cracks begin perpendicular to strands and end perpendicular to bottom of girder
Restraint by the bed
DEMEC strips
Figure 5.10 Typical cracking at end of Type IV girder after cut-down due to sliding friction
Camber measurements are shown in Figure 5.11. Internal strain measurements at midspan of the Type II and Type IV girders are shown in Figure 5.12.
4.0
3.5
Camber (in)
3.0
2.5
Girder 1.4
2.0
Girder 1.5
Girder 2.4
1.5
Girder 2.5
predicted, span 1
1.0
predicted, span 2
0.5
0.0
0
1
2
3
4
5
6
7
8
Days After Release
Figure 5.11 Camber-time measurements for each girder during the first 7 days
5-9
Vertical distance from bottom of beam (in) Vertical distance from bottom of beam (in)
60
After ES, calc
immediate, 1.4
immediate, 1.5
2 day, 1.4
50
2 day, 1.5
40
30
20
10
60
0
After ES, calc
-1200 -1000 -800 -600 -400 -200 0
immediate, 2.4
Total strain (microstrains)
immediate, 2.5
2 day, 2.4
50
2 day, 2.5
40
30
20
10
0 -1200 -1000 -800 -600 -400 -200 0
Total strain (microstrains)
Figure 5.12 Calculated and temperature corrected experimental strain profiles for Type II, span 1 girders (top), and Type IV, span 2 girders (bottom)
5-10
5.3 Transfer Length Each Type II and Type IV girder used 14 and 56 0.6-in., grade 270, low relaxation
prestressing strands, respectively. Transfer length of the 0.6-in. diameter prestressing strands was determined by taking concrete surface strain (CSS) readings at each end on each side of the four instrumented girders several times within the first 7 days after release. The 95% Average Mean Strain method developed by Russell (1992) was employed.
The average transfer length at the ends of the two Type II girders at 7 days after casting was 16.4 in.; this was 55% and 43% of the values found by AASHTO (16th edition) and ACI 318-99, respectively. This agreed well with the average of four previous tests done on Type II girders made with the grade 2 HPC field mix by Kahn, et al. (2002) of 17.0 in. at 7 days.
The average transfer length for the Type IV girders, however, was 27.3 in.. This seemed to imply that the transfer length was greater in a larger flange full of closely spaced strands than for a Type II flange with less strands. However, closer inspection of the strain data showed an offset between the end of the girder and the beginning of bond (Figure 5.13). It was concluded that this offset was caused by the vertical, sliding-friction cracking observed at the girder ends; the strand pulled through the cracked end of the girder, developing no bond stresses, then bonded to the girder for an additional distance beyond the restraint crack until the force was transferred. The "bond length" beyond the farthest crack of the Type IV girders was an average of 18.4 in. for the four Type IV girder tests. Figure 5.13 shows the concrete strain plotted as a function of the distance from the end of the girder. The initial linear strain trend intersects the 95% average mean strain plateau at 32.7 in.. If the cracking at the bottom of the girder ends was mitigated, the "bond length" described earlier would be the transfer length. Therefore, if the friction-restraint cracking were eliminated, the transfer length in the Type IV girder with 56 strands would be only 8% longer than in the Type II girders with just 8 strands.
In all cases, the transfer length was less than the 50 db (30 in.) given in ACI-318 and in AASHTO (16th Edition) and the 60 db (36 in.) given in AASHTO LRFD.
5-11
Strain (microstrains)
1800
1600 1400
Transfer Length = 32.7 in
1200 1000
800
Bond Length = 22.7 in
600
400
200
0
0
10
20
30
40
50
Distance from End of Beam (inches)
Figure 5.13 Transfer length plot at 7 days, Girder 2.5, east end (Slapkus and Kahn, 2002)
5.4 Deck Construction The HPC deck for Phase 1 was cast between December, 2001 and January, 2002. The
concrete used to cast the deck was specified as a "Class AA HPC" by the Georgia DOT. It was developed as a grade 1 HPC mix, intended for use in bridge decks in the laboratory study by Georgia Tech (Lai, et al., 1999). The mix design used by TCC constructors is given in Table 5.2
Each span of the deck was cast on a different day. Span 1 was cast first, followed by span 4, then span 2, and then span 3. The construction sequence for the deck is discussed chronologically.
After all of the diaphragms and edge beams were cast using a conventional concrete mix, galvanized metal deck forms were set in place. The stirrups protruding from the girders were bent to be parallel to the deck surface, approximately 5 in. clear of the top of the slab. The instrument leadwires were unbundled and extended along their respective girder lines to the west side of the bridge at span 1. The extension wires for the instruments to be placed in the deck were also run together with the wires from the girders. All wires were bundled
5-12
together and then tied to stirrups. Holes were drilled in the metal deck forming above the west abutment at Bent 1 and the wires were run through the decking.
Table 5.2 Class AA HPC Deck Mix Design
Component Type Cementitious material (lb/yd3): cement, Siam type I silica Fume, Euclid MSA Aggregates (lb/yd3): fines, Atlanta Sand Burke Pit coarse, Griffin, #67 granite Water (lb/yd3): Chemical Admixtures (oz/yd3): AEA, Euclid AEA 92 WRDA, Euclid WR-91 HRWR, Euclid Eucon 1037
TCC Mix Design
651 70
965 1837 225
19.5 143.2 16.2
Georgia DOT mix design tolerances
+/-1% -
+/-2% +/-2% +/-1%
-
Deck reinforcement was tied in place after the wires were run. The top and bottom steel mat was then tied. Epoxy coated bars were used for the steel mat closest to the deck surface, and normal mild steel reinforcing was used for the bottom mat. Bars were #5, spaced approximately 9 in. on center, both ways, top and bottom.
After all of the steel was tied in place, the researchers could install instruments that would be embedded in the deck. The vibrating wire strain gauges were installed at midspan and quarter-span directly above each instrumented girder. Additional gauges were set in place in the deck halfway between the centerlines of girders 4 and 5 at the midspan and quarter-span of span 1 and 2. Gauges were tied at the same elevation as the centerlines of the epoxy coated steel deck bars. Chemically inert composite reinforcing bars were field cut and used to position and secure the gauges to the deck bars. Figure 5.14 and 5.15 show installation of vibrating wire strain gauges in the deck.
5-13
Figure 5.14. Installing instrumentation in Phase 1 deck Thermocouple
VWSG
Figure 5.15. Close-up view of vibrating wire strain gauge (VWSG) and thermocouple in deck 5-14
Concrete placement for the deck took place for span 1 on November 16, 2001 and for span 2 on November 30, 2001. Spud vibrators were used on the freshly poured concrete, and a motorized rotating drum screed spanning the entire width of the bridge followed. Water mist was occasionally sprayed onto the freshly screeded concrete to prevent excessive drying. After the entire deck was cast, burlap was dragged across the entire span to roughen the surface. After initial set, tarps were placed over the deck for one day. After the deck had hardened, the contractor ground the deck to the proper elevation. The deck was then grooved perpendicular to the traffic flow. A permanent barrier was added to the north side of the bridge. Holes were drilled in the deck that were used to mount the deflection device inserts used for the load test. Stainless steel inserts were epoxied into the deck.
5.5 Material Evaluation 5.5.1 Girder Concrete
All concrete specimens were cast from the actual batches used to cast the girders and the deck. Each of the 8 batches used to cast the deck for span 1 and the 6 middle batches for span 2 had some type of specimens made.
Compressive strength was determined by testing 4" x 8" cylinders according to ASTM C 39. The chord modulus of elasticity was tested using 6" x 12" cylinders loaded in compression according to ASTM C 469. 4" x 4" x 14" beams were tested according to ASTM C78 to find the modulus of rupture. The beams were tested in four point bending. Creep was determined according to ASTM C 512. For both creep and shrinkage, 4" x 15" cylinders were cast horizontally with two sets of DEMEC gauge inserts spaced 10 in. apart on opposite sides. DEMEC inserts were connected to screws through holes drilled in the metal forms; after initial set of the concrete at about 4 to 6 hours, the screws holding the inserts were removed to allow free movement of the inserts during curing. Chloride ion penetration through the concrete was measured by performing a rapid chloride permeability test according to ASTM C 1202. 4" x 8" cylinders were cast for these tests. Just before testing, the 8" tall cylinders were cut into four 4" diameter x 2" thick cylinders with a diamond blade concrete cutting saw. The two outer 2" cylinders were discarded, and the two inner cylinders were tested. The coefficient of thermal expansion (CTE) was tested according to the Army Corps of Engineers Specification CRDC39.
5-15
The same type of 4" x 15" cylinders cast with a set of DEMEC gauge inserts used for measurement of creep and shrinkage strains were also used for CTE tests. An environmental chamber, Thermatron model SE-1200-3-3, was used to heat the specimens. Specimens were heated to 140 oF for 6 hours and four DEMEC concrete surface strain readings were taken. The specimen was then cooled to 40 oF for six hours and four readings were taken.
The direct pull out capacity of 0.6-in. prestressing strands was measured by embedding strands in a concrete block as described by Logan (1997). The block was 24" thick x 24" deep x 36" wide. Six 48 in. long prestressing strands were cut from the same reel as those used in the girders. These were embedded 20 in. into the block, with a 2 in. bond breaker at the top, for a total embedded length of 18 in..
Three different curing methods were used during the first 24 hours after concrete specimens were made: ASTM (ambient), insulated, and match-curing. All three methods were employed when curing the girder specimens, and only ASTM curing was used when curing the specimens made from the deck concrete. ASTM specimens were cured at ambient temperatures as described by ASTM C 31. They remained where they were cast for 6 hours, and were then placed in an open box in a truck until the next morning. Insulated cured cylinders remained where they were cast for 6 hours and then were placed in insulated boxes that allowed them to cure at a higher temperature caused only by their own heats of hydration until 24 hours after casting. Match-cure cylinders were cast and placed next to the girders. A thermocouple was embedded directly adjacent to thermocouple TC A.B cl 1 (middle of the bottom flange) and connected to a control panel that heated all of the matched-cure molds. The rates of hydration of the insulated cured specimens and those of the girder matched-cured cylinders compared reasonably well. The 56-day compressive strength of the match cured cylinders were within -3.2 to +2.8 % of the strength of the insulated cured cylinders for the span 1 (1S) and span 2 (2S) mixes.
The compressive strengths of the girder concrete are given in Table 5.3, and the strength gain is shown in Figure 5.16. Elastic modulus results are given in Table 5.4. Equation 5-1, found by Lai, et al. (1999) was the most accurate for predicting the elastic modulus, with an overall average overestimation of 56-day moduli of 3.4%.
5-16
Table 5.3 Compressive Strengths of Girder HPC
Span Identifier
1S 1S 1S 1S 1S 1S 1S 1S 1S 1S 1S 1S
Time of test
(days) 1 1 1 7 7 8 28 28 56 56 56 56
Curing Type
Insulated ASTM Match (SCP) Insulated ASTM ASTM (SCP) Insulated ASTM Insulated ASTM ASTM (SCP) Match
Average Strength
(psi) 10,434 7,700 10,819 11,055 10,641 12,647 12,276 12,435 12,435 13,379 15,228 12,795
Average Strength (MPa)
71.9 53.1 74.6 76.2 73.4 87.2 84.6 85.7 85.7 92.2 105.0 88.2
Number of Tests
18 3 1 3 3 1 3 3 3 18 1 6
Coeffiecient of variation,
% 5.2 6.5 n/a 2.0 3.6 n/a 2.5 6.1 6.1 3.6 n/a 3.5
2S
1
Insulated
7,401
51.0
32
16.1
2S
1
ASTM
6,343
43.7
6
9.6
2S
1 Match (SCP) 10,109
69.7
1
n/a
2S
7
Insulated 10,324
71.2
6
6.7
2S
7
ASTM
10,081
69.5
6
8.0
2S
8 ASTM (SCP) 10,892
75.1
1
n/a
2S
28
Insulated 11,928
82.2
6
7.7
2S
28
ASTM
12,527
86.4
6
8.1
2S
56
Insulated 12,453
85.9
6
9.5
2S
56
ASTM
13,157
90.7
32
6.6
2S
56 ASTM (SCP) 13,677
94.3
1
n/a
2S
56
Match
12,047
83.1
4
5.2
5-17
Cylinder Strength (psi) Cylinder Strength (psi)
16,000
16,000
14,000
14,000
12,000
12,000
10,000
10,000
8,000
6,000
4,000 0
Insulated cure Insulated Average ASTM Cure ASTM Average Match Cure, 56 days
10 20 30 40 50 60 Time After casting (days)
8,000
6,000
4,000 0
Insulated Cure Insulated Average ASTM Cure ASTM Average Match Cure, 56 days
10 20 30 40 50 60 Time after casting (Days)
Figure 5.16 Compressive strength gain for Span 1 (left) and Span 2 (right) girder HPC
Table 5.4. Elastic modulus results for concrete used to cast the girders
Span Identifier
1S 2S 1S 1S 2S 2S
Time of test
(days) 1 1 56 56 56 56
Curing Type
insulated insulated insulated ASTM insulated ASTM
Average Compressive Strength (psi)
Experimental modulus (ksi)
Number of Tests
Coefficient of variation,
%
10,434
3,436
3
3.5
7,401
3,397
6
0.4
12,435
5,031
3
1.3
13,379
4,983
6
2.2
12,453
4,911
6
4.3
13,157
4,962
6
1.9
5-18
where Ec = fc' = wc =
( ) Ec = 38,000
fc
'
+
730,000
wc 145
1.5
(psi)
(5-1)
modulus of elasticity (psi) concrete compressive strength (psi) concrete unit weight (lb/ft3) (average at 56 days was 146.9 lb/ft3)
At 24 hours, the average Poisson's ratio for the 9 specimens was 0.1672. At 56 days,
the average Poisson's ratio for 21 specimens was 0.1466.
The modulus of rupture is given in Table 5.5. If the average modulus of rupture, fr, is
represented by equation 5-2, then the constant would average 10.8 based on the data.
fr = fc'
(psi)
(5-2)
where,
= a constant
Table 5.5 Modulus of Rupture
Span Time of test Identifier (days)
1S
56
1S
56
2S
56
2S
56
Curing Type
insulated ASTM insulated ASTM
Average Strength (psi)
Experimental modulus of rupture (psi)
12,435
1,184
13,379
1,169
12,453
1,266
13,157
1,272
Number of Tests
3 5 13 3
The creep coefficient, defined as the creep strain divided by the elastic strain, was plotted versus time for the girder concrete mixes. The creep coefficient was predicted by Lai, et al. (1999) as well as Shams and Kahn (2000), then by ACI committee 209 (1997). The general form of the prediction of creep coefficient, t, at time t (in days) is given by Eq. 5-3.
5-19
where, d = t, = u =
t
=
d
t 0.6 +t
0.6
u
(unitless) (5-3)
days after load application when 50% of ultimate creep occurs creep coefficient at time t ultimate creep coefficient
Figure 5.17 compares the creep coefficient in this study to those other prediction methods. The ultimate creep coefficient for this study was 1.69. The base ultimate creep coefficient suggested by ACI 209 (1997) was 2.35; factors specific to this mix altered this value, including loading age, relative humidity, volume to surface ratio, slump, fine aggregate content, and air content, making the final ACI 209 predicted value 2.49. The ultimate creep coefficients found by Shams and Kahn (2000) and Lai, et al. (1999) were 1.23 and 1.42, respectively, for grade 2 HPC. More extensive data are given by Lopez and Kahn (2004).
2.0
1.6
Creep Coefficient, t
1.2
0.8
0.4
0.0 0
Measured Lai et al., 1999 Shams and Kahn, 2000 ACI-209, 1997
50
100
150
200
250
300
Age since loaded (days)
Figure 5.17 Creep coefficient for girder concrete compared to predicted values
5-20
The average shrinkage strains for the girder concrete are given in Figure 5.18 where they are compared with predicted values.
Microsstrains (in/in)
800 700 600 500 400 300 200 100
0 0
Measured Lai et al., 1999 Shams and Kahn, 2000 ACI-209, 1997
50
100
150
200
250
300
Age (days)
Figure 5.18 Shrinkage strains for girder concrete
Results of the chloride permeability tests are given in Table 5.6. The average total charge passed in each of the tests was 198.5 coulombs, which was less than one-tenth the maximum specified. The results from every test classified the chloride permeability as "very low" according to ASTM C 1202.
The coefficient of thermal expansion was measured for one specimen from the span 2 girders. Four measurements were taken at the high and low temperatures. The coefficient of thermal expansion for this test was 5.9125 /oF (10.643 /oC). These values were higher than the average values of 5.13 /oF (9.25 /oC) found by Shams and Kahn (2000) for 10,000 psi design strength HPC.
The direct pull-out capacity of the 0.6-in. strand was measured with strands embedded in a single block made from the concrete used to pour the second span. The average pull-out capacity of 59 kips shows that the stands were yielding considerably before pull-out occurred;
5-21
the corresponding bond stress was 1,304 psi. The yield force of the strand was 56.9 kips. The pull-out resistance in this test was similar to the average value of 56.3 kips found by Reutlinger (1999) for the Type II girder tests.
Table 5.6. Results of the chloride permeability tests for concrete used to cast the girders
Specimen Time of Identifier test (days)
2S
56
2S
56
2S
56
2S
56
1
28 +
2
28 +
Curing Type
ASTM ASTM ASTM ASTM
core core
Coulombs passed
225 222 174 170 200 200
Permeability
very low very low very low very low very low very low
5.5.2 Deck Concrete The Class AA HPC concrete used to pour the deck had a minimum required strength of
7,250 psi (50 MPa) at 56 days. The results of compression tests for the concrete used to cast the deck for spans 1 and 2 are given in Table 5.7. Figure 5.19 show graphs of the concrete strength vs. time for the deck, spans 1 and 2.
Table 5.7 Deck concrete strength results (D1 is for span 1 and D2 is for span 2)
Span Time of test Curing Type Identifier (days)
D1
1
D2
1
D1
3
D2
3
D1
7
D2
7
DOT-1
7
DOT-2
7
D1
28
D2
28
D1
56
D2
56
DOT-1
56
DOT-2
56
ASTM ASTM ASTM ASTM ASTM ASTM ASTM ASTM ASTM ASTM ASTM ASTM ASTM ASTM
Average Strength
(psi) 2,073 2,497 3,922 4,411 4,813 5,268 5,845 6,287 5,719 6,715 6,230 6,880 7,370 9,329
Average Strength (MPa)
14.3 17.2 27.0 30.4 33.2 36.3 40.3 43.4 39.4 46.3 43.0 47.4 50.8 64.3
Coefficient
Number of of variance,
Cylinders
%
8
20.5
6
9.5
8
13.5
6
7.1
8
9.5
6
7.7
6
11.3
6
6.9
8
10.5
6
9.6
8
10.7
6
7.2
6
7.0
8
7.9
5-22
Compressive Strength (psi) Compressive Strength (psi)
8,000 7,000 6,000 5,000 4,000 3,000 2,000
9,000 8,000 7,000 6,000 5,000 4,000 3,000 2,000
1,000 0
1,000
10
20
30
40
50
60
0
Cylinder age (days)
10
20
30
40
50
60
Cylinder age (days)
Figure 5.19 Deck concrete strength gain over time (span 1 left and span 2 right)
The average 56-day modulus of elasticity for the deck span 1 concrete was 3,546 ksi, and that for span 2 was 3,673 ksi. These values were 5% less than calculated by Equation 5-1, above.
Figure 5.20 compares the shrinkage strains in this study to other prediction methods. The ultimate shrinkage strain predicted for a 20 year period for this study was 676 microstrains. The base ultimate shrinkage strain suggested by ACI 209 (1997) was 780 microstrains.
The permeability requirements for both the Class AA concrete and the grade 1 HPC were that the total coulombs passed in a chloride permeability test be less than 2,000 coulombs. Four tests were done for each span at 56 days for the concrete used to make the deck. Four additional tests were performed by the Georgia DOT at 56 days for span 2 only. The average total charge passed in the tests was approximately twice the maximum charge allowed. The average for span 1 was 5,047 coulombs, and the average for span 2 was 4,320 coulombs. The results from 8 of the 12 tests classified the chloride permeability as "high".
5-23
Microstrains (in/in)
800 700 600 500 400 300 200 100
0 0
Measured Lai et al., 1999 Shams and Kahn, 2000 ACI-209, 1997
50
100
150
200
Drying Time (days)
Figure 5.20 Average shrinkage strains vs. time for the two deck specimens
The coefficient of thermal expansion for the deck concrete was 6.3500 /oF (11.430 /oC). The average unit weight found at 56 days for two tests of the span 1 mix was 145.1 lb/ft3. The average unit weight of the span 2 mix was 144.8 lb/ft3.
5.6 Load Test on Phase 1 Static live load tests were performed on the bridge using two fully loaded dump trucks.
The first test took place on January 28, 2002 between 5:30 and 7:00 AM. The axle layout and load distribution for each truck used during the load tests is given in Figure 5.21. The gross vehicle weight of each of the trucks was approximately 56 kips. Figure 5.22 is an overview plan of spans 1 and 2.
5-24
Front axle weight: 14.95 kips
82"
Truck 1 Tag # 4069520 GVW: 56.34 kips
184"
Front rear axle weight: 20.84 kips
52"
back rear axle weight: 20.55 kips
Front axle weight: 16.44 kips
82"
Truck 2 Tag # 4069530 GVW: 56.37 kips
184"
Front rear axle weight: 20.23 kips
52"
back rear axle weight: 20.07 kips
Figure 15.21 Axle layout and wheel load distribution for static live load tests trucks
BENT 3
2.6 2.5 2.4 2.3 2.2 2.1
cl
SPAN 2
North
qp BENT 2
1.6 1.5 SPAN 1
1.3 1.4cl
BENT 1
1.2
qp
1.1
Figure 5.22 Overview plan of spans 1 and 2 used to diagram live load tests 5-25
In order to create maximum point loads near the midspan and quarterspan of the instrumented girders during the load tests, the vertical centroid of the two rear axles of each truck were positioned as close as possible these points. Trucks were positioned side by side for load cases 1 through 4. For load cases 5 through 7, they were positioned one in front of the other in a "train" configuration. Truck number 1 was always positioned horizontally on the diagrams with its wheels straddling girder 5. Truck number 2 was positioned to the left of truck number 1, straddling girder 4, for load cases 1 through 4 and behind truck number 1, straddling girder 5, for load cases 5 through 7. As an example, load case 4 and 7 are illustrated in Figures 5.23 and 5.24, respectively.
24"
SPAN 2
2.1 2.2 2.3 2.4 2.5 2.6
Figure 5.23 Load case 4 with trucks side-by-side and rear axels at midspan of span 2 Figure 5.25 is a plot of the measured and predicted strain profiles for load case 7, span 2. The non-linearity in the deck strains is shown on this plot; this was the most non-linear plot of all cases and is believed to have occurred because of the torsion in the girders due to the skew in the bridge. The larger deck strains were in girder 2.5 relative to 2.4, and the intermediate deck strain was halfway between the two points. The finite element analysis gave a close approximation to the experimental data.
5-26
2.1 2.2 2.3 2.4 2.5 2.6
28" 7"
SPAN 2
27"
Vertical distance from bottom of beam (in)
Figure 5.24 Load case 7 with trucks in a train on span 2
75
2.4 cl, FEA
2.5 cl, FEA
2.4 cl, exp
60
2.5 cl, exp
2 d cl, exp
45
30
15
0
-80
-60
-40
-20
0
20
40
60
80
live load strain (microstrains)
Figure 5.25 Plot of the measured and predicted strain profiles for load case 7, span 2
5-27
Deflections for Load Test 1 are given in Table 5.8 and are compared to finite element analysis results and to deflections calculated based upon strain profiles.
Table 5.8 Deflection results for Load Test 1
Load Case
LC 1 LC 2 LC 3 LC 4 LC 5 LC 6 LC 7
FEA Span
position
1, diaphragms
Span 1, measured
included
1.4 cl 1.5 cl 1.4 cl 1.5 cl 2.4 cl 2.5 cl 2.4 cl 2.5 cl 1.4 cl 1.5 cl A.4 cl A.5 cl 2.4 cl 2.5 cl
-0.1088 -0.0926 -0.1355 -0.1336
-0.0625 -0.2500 -0.0625 -0.3125
-0.0723 -0.0966 -0.0513 -0.0726
-0.0625 -0.0625 -0.0625
-0.0156
Deflections, inches
Span 1, calc'd from internal strains
FEA Span
2,
Span 2,
diaphragms measured
included
-0.1000 -0.1000 -0.0700 -0.0800
-0.3084 -0.3143 -0.3830
-0.2031 -0.4063 -0.2344
-0.0500 -0.0800 -0.0300 -0.0500
-0.3986
-0.1945 -0.2316
-0.4688 -0.1563
-0.3530 -0.0938 -0.4166 -0.2813
Span 2, calc'd from internal strains
-0.1000 -0.0700 -0.2400 -0.2800
-0.1200 -0.1600 -0.2700 -0.3700
5.7 Long-term Bridge Performance Figures 5.26 and 5.27 show the exact locations of the VWSG in span 2 and span 1 girders
and deck. Figure 5.28 presents the strain profiles of Girder 2.4 at different times: immediately after strand release, before deck placement, after deck placement, 1, 2, and 3.2 years after strand release.
5-28
Span 2 gages, at the centerline only, in the
girders
Span 2 gages at the quarterpoint only, in the
girders
VWSG 2.B cl 3
VWSG 2.B qp 3
Span 2 gages in the deck above the girders
3"
54"
VWSG 2.B cl 2
VWSG 2.B qp 2 38.5"
10"
23.3" 25.2"
VWSG 2.B xx 4
VWSG 2.B cl 1
3"
VWSG
3"
2.B qp 1
Figure 5.26 Cross section locations of the vibrating wire strain gauges in span 2
Span 1 gages in the girders
36" VWSG 1.B xx 2
VWSG 1.B xx 1
VWSG 1.B xx 3
26.5"
4" 3"
Span 1 gages in the deck above the girders
3"
VWSG 2.B xx 4
Figure 5.27. Cross sectional locations of the vibrating wire strain gauges in span 1 5-29
After Strand Release
After 1 year
After 80days / Before Deck Placement
After 2 years
After 80days / After Deck Placement
After 3.2 years
30
Distance from Cross-section CG (in)
20
10
0
-1050 -850
-650
-450
-250
-50
-10
-20
-30 Total Strain ()
Figure 5.28 Strain profiles versus time of Girder 2.4
Several comparisons can be made from Figure 5.28. The profiles after strand release and before deck placement clearly show the time dependent, creep and shrinkage effects. Immediately after release, the strain profile shows compression in the bottom of the girder close to 1,000 , but 80 days later that strain had decreased to 300 . The losses are larger for the bottom of the girder where more prestress was applied. After the deck placement the profile is inverted, showing the maximum compressive strain at the top of the girder. This large change in strains is due to the superimposed dead load moment. The experimental strain profiles due to strand release, and deck placement were similar to those calculated from a finite element model (Slapkus and Kahn, 2002).
The strain profiles after one, two and three years changed little in comparison with the profile after the deck placement. Moreover, there was almost no change in curvature which indicates that there was no increase in deflection. The change in compressive strain at the top of the girder between 80 days (deck placement) and 1,200 days (3.2 years) was 131 while that change at the top row of the bottom strands was 137 .
5-30
Figure 5.29 presents the axial strain versus time recorded in the top surface of the girder and in the deck at the midspan of Girder 2.10. Figure 5.28 also shows the strain profiles obtained 4, 12, 17, and 23 months after deck placement.
Microstrains (in/in x 10-6)
300
2.10.CL.top
2.10.CL.mid
2.10.CL.girder
200
100
0
-100 0
-200
-300
-400
-500 0
100
200
300
400
500
600
700
800
Time after deck placement (days)
t
@68oF
@65oF c
c
@64oF c @67oF
Deck c t
ct
ct
ct
Girder
Measured Profile Assumed Profile
Figure 5.29 Strain versus time at midspan of Girder 2.10
As shown in Figure 5.29, the most changes occurred during the first 100 days after deck placements. The electrical resistance strain gauge (ESG) attached to Girder 2.10 showed that the top of the girder underwent a compressive strain of 100 immediately after deck placement, and the strain increased to 300 approximately at 100 days. Initially the girder showed an elastic deformation due to the positive moment imposed by the self weight of the deck. However, the compressive strain at the top of the girder increased during the first three months.
5-31
This is believed to be an effect of early thermal contraction and of the drying shrinkage of the deck. The girder restrained the shrinkage of the deck and, thereby, imposed a tensile strain in the deck which is shown in the 3-month profile shown in Figure 5.29.
Between 100 days and two years after the deck placement, the axial strain in the three ESG located at the midspan of the girder showed small changes and maintained the relative difference between each other. This is shown by the little difference between the strain profiles after 12, 17, and 23 months. As seen for Phase 1, the strain profile did not change the curvature leading to the conclusion that after 100 days there were no further changes in deflection. The west end of girder 2.10 was instrumented in the same way as the midspan. The observations made for the midspan analysis, are entirely applicable to the data recorded for the end of the girder.
Figures 5.30 and 5.31 present the deflection in the bridge and the temperature after deck placement in linear and logarithmic scale, respectively. Figures 5.30a and 5.31a present the deflection at the midspan of girders 2.9, 2.10 and 2.11. A negative change in deflection represents a downwards displacement. Figures 5.30b and 5.31b present the ambient temperature and the temperature in the deck for each of the deflection measurements.
The placement of the deck caused a downward deflection at the midspan of 2.67 in. A few hours after casting, the average temperature in the deck rose from 76oF to 108oF due to the heat of hydration, even though the ambient temperature dropped from 63oF to 55oF. This change in temperature heated the top of the girders producing an upward change in deflection of 0.43 in. During cooling of the deck, the bridge had a downwards deflection of 0.76 in. The resulting total deflection two days after deck placement was 3.00 in. Therefore, between zero and two days after the deck placement, the deflection increased 0.33 in. due to deck contraction. During the next 14 days, the deflection increased 0.25 in. more. One month after the deck placement, the south concrete barrier of the bridge was cast. This structure imposed an additional self weight that, unlike the deck, was not evenly distributed among the girders.
One week after the barrier placement, the two phases of the bridge were connected by casting a longitudinal joint between the two decks. After the two phases were connected, Phase 1, built one year before, restrained the downward displacement of the north edge of the Phase 2 deck. Therefore, the weight of the barrier on the south edge of the deck, the restraint from Phase 1 on the north edge of the deck and the torsional effects produced by the skew in the bridge made
5-32
Change in deflection (in)
Temperature (oF)
a 0.00 -0.25 -0.50 -0.75 -1.00 -1.25 -1.50 -1.75 -2.00 -2.25 -2.50 -2.75 -3.00 -3.25 -3.50 -3.75 0
120
b 110 100
90
80
70
60
50
40
30
20
10
0 0
Initial deflection due to self weight of the deck
Barrier placement
Connection with Phase 1
North Pin Center Pin South Pin Average
100 Winter 02/03
200
300
400
500
600
700
Time after deck placement (days)
Deck average Temp(F) Ambient Temp (oF) Winter 03/04
Summer 04
Summer 03
100
200
300
400
500
600
700
Time after deck placement (days)
Figure 5.30 (a) Deflection in girders 2.9, 2.10, and 2.11 versus time in linear scale. (b) Ambient temperature and deck temperature versus time in linear scale.
the future deflection behavior of the three girders vary. The girder closest to the north edge (2.9) decreased its deflection from 3.15 to 2.75 in while Girder 2.11 (the closest to the south edge) increased the deflection to 3.62 in. The only noticeable changes after 30 days in the center girder (2.10) and in the average deflection for all three girders were due to seasonal changes. An increase in ambient temperature produced an upward movement of the bridge at the midspan while a decrease in temperature produced a downward movement in the bridge. The deflection
5-33
at the midspan of Span 2, Phase 2 varied between 3.00 and 3.25 in. due to temperature changes without showing any further increase in creep and shrinkage related deflection after 30 days.
Change in deflection (in)
a
0.00 -0.25
-0.50
-0.75
-1.00
-1.25
-1.50
-1.75
-2.00
-2.25
-2.50
-2.75
-3.00
-3.25
-3.50
-3.75 0.1
Temperature (oF)
120
b 110
100 90 80 70 60 50 40 30 20 10 0
0.1
Initial deflection due to self weight of the deck
Connection with Phase 1
Barrier placement
North Pin Center Pin South Pin Average
1
10
100
Time after deck placement logarithmic scale (days)
1000
Deck average Temperature (oF) Winter 03/04 Ambient Temp (oF)
Winter 02/03
Summer 04
Summer 03
1
10
100
Time after deck placement logarithmic scale (days)
1000
Figure 5.31 (a) Deflection in girders 2.9, 2.10, and 2.11 versus time in logarithmic scale. (b) Ambient temperature and deck temperature versus time in logarithmic scale.
Figure 5.32 shows the strain profile obtained at the midspan of Girder 2.4 for the seasonal measurements. The thermal variation was obtained as the difference between the total strain measured by the strain gauge and the creep and shrinkage related strain.
5-34
Distance from girder cross-section CG (in)
50 40 30 20 10
0 -10 -20 -30
-80
Winter 2001-02
Summer 2002 Winter 2002-03 Summer 2003
Winter 2003-04
Summer 2004
Winter 2004-05
-60
-40
-20
0
Total Strain ()
Figure 5.32 Seasonal strain profile change at midspan Girder 2.4
Figure 5.32 shows that after three years of measurements the seasonal variations account for approximately 40 of lateral shortening/expansion. This 40 is more than the long-term creep and shrinkage change between 2 and 3.2 years. The change in the strain profile angle between winter and summer is due to a thermal gradient from bottom of the girder to the top of the deck along with the thermal mismatch between deck and girder. This angle variation represents a change in curvature which results in a change in camber/deflection.
Figure 5.33 presents the deflection recorded for Phase 2, for the seasonal variation and for one morning/afternoon variation. The maximum seasonal variation recorded for was between winter 2002-03 and summer 2004; the change was 0.25 in. This seasonal change represents roughly one-third of the time-dependent load related deflection.
5-35
Deflection (in)
-4.0
North Pin Center Pin South Pin Average
-3.5
-3.0
-2.5
-2.0
-1.5
-1.0
-0.5
0.0 Winter Summer Winter Summer 2002-03 2003 2003-04 2004
May 16th 2003 8:00 AM 5:00 PM
Figure 5.33 Seasonal and daily deflection at midspan Girder 2.10
Figure 5.33 also shows the change in deflection which occurred between 8:00 AM and 5:00 PM on May 16th, 2003 approximately six month after the deck placement. The one-day change in deflection was 0.31 in. This one day change in deflection was as great as the seasonal variation. Further, this deflection due to thermal variations was about the same as the deflection due to loading span 2 of Phase 1 with the two fully loaded dump trucks.
5-36
Chapter 6. Type IV and Bulb-T 56 Flexural Test Results
6.1 Introduction The specific objectives of this study were as follows: Construct two composite girders, an AASHTO Type IV and PCI modified Bulb-tee 56, with the same mix design and approximately the same strand arrangement as that used in the Jonesboro Road demonstration bridge girders discussed in Chapter 5. Perform a static load test on each girder to determine the actual flexural strength in order to compare the actual, experimental strength to those strengths predicted by the AASHTO Standard Specifications for Highway Bridges. Determine the transfer lengths of each girder to compare with those of Type II girders tested by Reutlinger, et al. (2002) and the Type IV girders reported in Chapters 4 and 5. Determine prestress losses for each girder to compare with values predicted by AASHTO. Further study the effect of composite deck contraction on the deflection of the girder.
6.2 Girder Design Duplicating the length and strand arrangement of the 124-ft girders of the Jonesboro
Road Bridge was not possible due to constraints of the structures laboratory at the Georgia Institute of Technology. As dictated by these constraints, a length of approximately 90-ft and a deck thickness of 8-in. were chosen as controlling parameters.
Although the exact strand arrangement could not be duplicated due to the girder length change, the objective was to have a flexural capacity as close to that of the bridge girders. Shear reinforcement identical to that used in the demonstration bridge would be used for half the span; shear reinforcement as designed by the GDOT computer design program would be used for the other half span.
The mix design and specified design compressive strength for each girder were to be the same as that used in the demonstration bridge (Chapter 5). The mix was classified as a "Class AAA HPC" by the Georgia DOT, and was developed as a grade 2 HPC. The mix
6-1
proportions of the concrete used are given in Table 6.1. The specified 56-day design strength was 10,150 psi (70 MPa).
Table 6.1 Mix design for Type IV and BT-56 girders
Mix Design for Bridge Girders & for Laboratory Girders
Component cement, type III (lb/yd3) Flyash, type F (lb/yd3) Silica Fume, Force 10,000 (lb/yd3) Fines, Brown Brothers #2 sand (lb/yd3) Coarse, Vulcan #67 stone (lb/yd3) Water (lb/yd3)
Water/Cementious Ratio AEA, Daravair 1000 (oz/yd3) Water reducer, WRDA 35 (oz/yd3) HRWR, Adva 100 (oz/yd3)
Retarder
Quantity 796 98 70
965 1837 237 0.25
7 35 169 0
The bridge construction used Type I cement, but Standard Concrete Products Company (SCP) changed to using only Type III cement in their mixes, therefore, Type III was used for the actual mix design for the Type IV and the BT-56 girders.
The GDOT in house beam design program "BRPSBM1" was used to design the girders so that the number of 0.6-in. diameterstrands could be maximized for a 90-ft long girder. The final designs for each laboratory test girder are shown in Figures 6.1 through Figure 6.3.
6-2
2.50"
51.5"
47.0"
5 SPACES @ 2.0"
2.75"
11 SPACES AT 2"
SECTION AT MIDSPAN
SECTION AT END
TYPE IV STRAND ARRANGEMENT
6.0" 2.50"
53.5" 40"
4 SPACES @ 2.0"
2.75"
11 SPACES AT 2"
SECTION AT MIDSPAN
SECTION AT END
BT-56 STRAND ARRANGEMENT
Figure 6.1 Cross section showing strand arrangements at midspan and end of girders
6-3
Figure 6.2 Type IV girder strand arrangement and shear reinforcement details 6-4
4 Spaces at 3.5"
2"
#6 U stirrup at each location
4 Spaces at 11.5"
Beam length = 89'-2"
Midpoint, details symmetrical
about midpoint unless noted
15 Spaces at 24.0"
6.5"
15 Spaces at 17.5"
#5 U stirrup at each location
42 Spaces at 8.5"
4"
4 Spaces at 4.0" 2"
#6 U stirrup at each location
2" Typ. 12"
54"
54"
CL Low friction type hold down
CL Bearing to CL Bearing = 88'
IV GIRDER ELEVATION
NOT TO SCALE
12"
# 3
23"
5"
# 6
# 5
350
58.5"
8"
13"
13"
650
550
78"
451
# 4
18" 452
# 4
17"
78" 450
IV REINFORCEMENT DETAILS
CL Bearing
7" Typ.
6-451
4"
450
4 Spaces at 4.0" 2"
#6 U stirrup at each location
8-350 AT 12" CL Bearing
2 5/8" 2-452 7" Typ.
IV GIRDER END REINFORCEMENT DETAIL
Figure 6.3 BT-56 girder strand arrangement and shear reinforcement details 6-5
4 Spaces at 3.5"
2"
#6 U stirrup at each location
5 Spaces at 9.5"
Beam length = 89'-2"
Midpoint, details symmetrical
about midpoint unless noted 16.5" 17 Spaces at 24.0"
12 Spaces at 17.5"
#5 U stirrup at each location
50 Spaces at 7.0"
4"
2-452
4 Spaces at 4.0" 2.0"
#6 U stirrup at each location
2" Typ.
12" # 6
13" 650
54"
54"
CL Low friction type hold down CL Bearing to CL Bearing = 88'
BT GIRDER ELEVATION
CL Bearing
7" Typ.
12" # 3
# 5 61"
23" 350 3.5"
102"
451
# 3
13"
38"
550
352
Beam length- 4" (24" Min. splice)
452
BT REINFORCEMENT DETAILS
# 4 8"
18" 450
# 4
2-452
4" 352
4 Spaces at 4.0"
2.0"
#6 U stirrup at each location
8-451 8-350 AT 12"
CL Bearing
2 5/8" 2-450 7" Typ.
BT GIRDER END REINFORCEMENT DETAIL
6.3 Instrumentation Various instrumentations were used in this study. Load cells, manufactured at Georgia
Tech, were used during girder construction to measure prestressing strand forces. Deflection measurements from casting up to the time of testing were made using a piano wire hung along side each girder, anchored at one end, and strung over a pulley with a weight attached at the other end.
Detachable mechanical strain gauge (DEMEC) inserts, spaced 2-in. apart, were located at the level of the level of top and bottom prestressing strands at each girder end and at midspan on both sides of the girder as shown in Figure 6.4. These DEMEC points were used to measure the transfer length at girder ends, whereas the DEMEC inserts at midspan were used to obtain strain profiles to double check strain profiles obtained from other strain gauges. Vibrating wire strain gauges (VWSG) with a 6-in. gauge length and having an attached thermistor, were embedded within both the girder and deck to obtain strain profiles at midspan and quarter-span. The labeling for the VWSG's was simply a "Q" or "M" for quarter-span or midspan, with a gauge number, starting at the section top and continuing down. In addition, thermocouples were embedded to serve as a back up for the thermistor readings; the latter were used for strain temperature corrections. The VWSG and thermocouple locations are shown in Figure 6.5, where "TC" indicates thermocouple. A third type of strain gauge used was an electrical resistance strain gauge. These gauges were epoxy mounted to the top surface of each girder prior to casting of the deck to provide a strain measurement at the girder-deck interface.
During load testing, other instruments were added in addition to those mentioned above. Additional electrical resistance strain gauges were surface mounted to each deck at midspan to measure extreme fiber compressive strains during testing. String potentiometers were attached to each girder bottom at midspan to measure girder deflection during testing. Other string potentiometers were mounted horizontally to each side of the girder at midspan, at the level of the bottom layer of prestressing strands to ultimately determine strains in the bottom layer of strands.
6-6
BEAM LENGTH = 89'-2"
24" DEMEC STRIP AT GIRDER CENTERLINE, BOTTOM AND TOP
2.50"
3.00"
TYPE IV GIRDER ELEVATION
30" DEMEC STRIP AT GIRDER TOP ONLY, BOTH ENDS
BEAM LENGTH = 89'-2"
24" DEMEC STRIP AT GIRDER CENTERLINE, BOTTOM AND TOP
18" HOLES FOR ATTACHING DEMEC
STRIPS TO FORMS
3.00"
2.50"
60" DEMEC STRIP AT GIRDER BOTTOM ONLY, BOTH ENDS
2.50" BT-56 GIRDER ELEVATION
18" HOLES FOR ATTACHING DEMEC
STRIPS TO FORMS
2.50"
Figure 6.4 Elevation of each girder with locations of DEMEC inserts
6-7
4.12"
M1 & TC1 2.625" M4
M2 & TC2 M3
7.50"
4.19"
3.87"
Q1
2.50"
Q4
Q2 Q3
7.62"
4.00" 8.00"
TC3
NORTH
TC4 M6
13" TC5
M5 51"
SOUTH
10.00"
M7 7.00"
2.0" TYP. Q5
NORTH Q6
2.75" Q7
51" SOUTH
27.75"
SECTION AT MIDSPAN
SECTION AT QUARTERPOINT
TYPE IV INSTRUMENTATION LOCATIONS
3.75"
M1 & TC1 3.00"
M4
M2 & TC2 M3
7.25"
4.50"
3.88"
Q1 2.76" Q4
Q2 Q3
7.50"
4.13"
3.375"
TC3
TYP.
NORTH TC4
M6
11" TC5
MM75 SOUTH 52.75"
M7 2.75"
SECTION AT MIDSPAN
Q5 NORTH
Q6
SOUTH 53"
23"
Q7 SECTION AT QUARTERPOINT
BT-56 INSTRUMENTATION LOCATIONS
Figure 6.5 Cross-sections at midspan and quarter-span for both girders showing instrumentation
6-8
6.4 Girder Construction The Type IV and BT-56 girders were cast at Standard Concrete Products (SCP) on
April 22nd and 23rd, 2004, respectively. The mix design for both girders was to match the concrete of the demonstration bridge discussed in Chapter 5, and the same 0.6-in. diameter, low relaxation prestressing strands were used. The actual mix design used is shown in Table 6.1. It was decided by SCP to construct both girders on the same prestressing bed so as not to slow down production, which is the reason for the 1-day separation between casting the two girders.
Load cells were installed at the dead end of the prestressing bed, as the prestressing strands were installed. Eight load cells were used to check the values of the prestressing force determined by SCP's instrument. The load cells were installed in the arrangement shown in Figure 6.6. Also in Figure 6.6 are the 10 strands not common to both girders. These 10 strands are marked with an "x" in the figure. Due to the different prestressing strand arrangement, the Type IV girder, consisting of 52 prestressing strands, was constructed first. The day after casting the Type IV girder, after the appropriate fci' was reached, the forms were removed and the 10 strands mentioned above were cut. Following this, the two, outer top strands in the BT56 girder were placed and stressed. Then the same casting process was repeated for the BT-56 girder. Figure 6.7 shows the BT-56 girder being cast.
On Saturday, April 24th, after a specified fci' for the BT-56 girder was obtained, the prestressing strands were cut. Both girders were moved off the prestressing bed that day, and soon thereafter were moved to a storage location on site. The Type IV girder was stored for a period of 18 days before being transported to the Georgia Tech Structures Laboratory on May 10, 2004. SCP allowed the BT-56 girder to be stored on site for a period of 71 days before being moved to the Structures Laboratory on August 3, 2004.
6-9
TYPE IV END
BT-56 END
Figure 6.6 Girder sections showing load cell locations and strands to be cut before the BT-56 girder pour
Figure 6.7 Tuckerbilt truck placing concrete into the BT-56 girder forms 6-10
The composite deck for both girders was cast in the Georgia Tech Structures Laboratory. The deck reinforcement details typical of both decks is shown in Figure 6.8. The formwork was designed and built to distribute the deck dead load to the girder, as is done in the field, with no shoring provided. The formwork detail and casting of the Type IV deck are shown in Figure 6.9. Each deck was water cured for 5 days, and the forms were removed seven days after casting.
#4 BAR @
10" O.C.
1'-4"
1.0"
#4 BAR @ 10' O.C.
#4 BAR, SPACING VARIES
1'-4" 2.75"
7.5" 7.5" 3.5"
Figure 6.8 Cross-section showing typical deck reinforcement for both girders
6-11
Figure 6.9 Screeding of Type IV deck in GT Structures Lab
6.5 Material Properties
Although each girder had the same mix design, the curing used on each differed. The Type IV girder was steam and water cured, while the BT-56 girder was cured without steam. For this reason, many of the reported concrete properties are kept separate and have individual averages for each girder. In addition, due to time constraints, some testing immediately after casting of the girders was done simultaneously for both girders. As such, the time of initial cylinder testing was at 24 hours for the BT-56 girder but at 48 hours for the Type IV girder, due to the 1-day casting difference. A few ASTM cured cylinders were tested at 24 hours for the Type IV girder, but the majority were tested at the "initial" 48-hour time. Girder and deck concrete properties are given in Tables 6.2 and 6.3, and the description of each different curing method used in the tables is as follows:
Sure Cure: Curebox: ASTM:
Cured in SCP's match curing apparatus and tested at approximately 14hours. Cured in insulated boxes on site, demolded and stored in the fog room until testing. Cured in ambient conditions on site, demolded and stored in the fog room until testing.
6-12
Table 6.2 Girder concrete properties
Girder
fci'
fc' ~14 Hrs (Interpolated Sure Cure from sure cure
fc' 24-hour
ASTM (psi)
(psi)
& curebox
data) (psi)
fc' 24-hour Curebox
(psi)
Type IV 10526
11413
10416
NA
Range NA
9007 -
NA
11458
fc' 48hour ASTM (psi)
10660 10189 10998
fc' 48hour Curebox (psi)
fc' 56-day Curebox
(psi)
fc' 56-day fc' Test-day
ASTM Curebox
(psi)
(psi)
11727 10059 13181
13660 13387 13915
15287 13714 16436
14353 12842 15376
BT 56 Range
9171 NA
9850
8986
11755
NA
10749 -
NA
8897 - 9068 12810
NA 13819 14841 15145 13334 - 14034 - 14626 14427 15965 15840
Table 6.2 Continued
Eci 24-hour Eci 48-hour Ec 56-day Girder Curebox Curebox Curebox
Ec 56-day
Ec Test-day fr 56 day Curebox ASTM
CTE 3-day
CTE 63-day
(psi)
(psi)
(psi) ASTM (psi) (psi)
(psi) (/C) (/C)
Type IV NA
4.67E+06 4.72E+06 5.11E+06
NA
Range
4.66E+06 - 4.63E+06 - 4.89E+06 4.69E+06 4.82E+06 5.31E+06
974
NA 10.92
950 -
1009
NA
BT 56 4.77E+06 Range 4.58E+06 -
4.88E+06
4.76E+06 5.12E+06 4.29E+06 4.60E+06 - 4.39E+06 - 4.23E+06 4.98E+06 5.15E+06 4.39E+06
1011 1001 1025
10.48 8.57 11.83
14.16 NA
Table 6.3 Deck concrete properties
Deck
fc' 24-hour fc' 56-day ASTM
ASTM (psi)
(psi)
fc' Test-day ASTM (psi)
Ec 24-hour ASTM (psi)
Ec 56-day
Ec
Test-day
CTE day
4- CTE 62-day
ASTM (psi) ASTM (psi) (/C) (/C)
Type IV 2977
7166
Range 2204 - 3926 6693 - 7844
BT 56
2352
6653
Range 1966 - 2770 5679 - 7585
7879 2.57E+06 3.56E+06 3.55E+06 8.33 9.84
2.05E+06 - 3.31E+06 - 3.36E+06 -
9.65-
7457 - 8282 3.09E+06 3.74E+06 3.63E+06 8.2 - 8.47 10.04
CTE 2- CTE day 56-day
(/C) (/C)
6701.30 2.47E+06 3.51E+06 3.51E+06 9.47 8.85
2.31E+06 - 3.34E+06 - 3.34E+06 - 9.28 - 8.53 6286 - 7238 2.70E+06 3.73E+06 3.73E+06 9.66 9.16
6-13
Figures 6.10 and 6.11 show concrete compressive strength versus time for each girder and for each deck. For each girder, a strength at time of release was interpolated from the SureCure strengths obtained by SCP and the first curebox strengths.
The coefficient of thermal expansion (CTE) was measured for both girders and decks, using 4-in. by 15-in. cylinders. Figure 6.12 plots show the average CTE for both girders and both decks and a best fit line for each plot. An average CTE value for both girders as well as the decks was found as there was no evidence that each batch of concrete should be treated differently. The best fit CTE over time was used in temperature compensation calculations so as to represent the CTE.
6-14
Compressive Strength (psi)
Type IV Girder Concrete Compressive Strength vs. Time
16,000
15,000
14,000
13,000
12,000
11,000 10,000
11,413 psi
9,000
8,000 0
10
20
30
40
50
60
Time After Casting (days)
Insulated Average
ASTM Average
Compressive Strength (psi)
BT-56 Girder Concrete Compressive Strength vs. Time
16,000
15,000
14,000
13,000
12,000
11,000
10,000 9,000
9,850 psi
8,000 0
10
20
30
40
50
60
Time After Casting (days)
Insulated Average
ASTM Average
Figure 6.10 Compressive concrete strength vs time for each girder
6-15
Compressive Strength (psi)
9000 8000 7000 6000 5000 4000 3000 2000 1000
0 0
Type IV Deck Concrete Compressive Strength vs. Time
10
20
30
40
50
60
70
80
90
100
Time After Casting (days)
Batch 1 ASTM Cured
Batch 2 ASTM Cured
Compressive Strength (psi)
BT-56 Deck Concrete Compressive Strength vs. Time
9000 8000 7000 6000 5000 4000 3000 2000 1000
0 0
10
20
30
40
50
60
70
80
90
100
Time After Casting (days)
Batch 1 ASTM Cured
Batch 2 ASTM Cured
Figure 6.11 Compressive concrete strength vs time for each deck
6-16
CTE (micro strain/degree C)
14
12
10 y = 9.5724x0.0683
8
6
4
2
0
0
25
50
75
100
125
150
175
200
225
Time from Casting (days)
Avg. CTE
Power (Avg. CTE)
Figure 6.12a. Plots of average CTE for both girders vs time and a trend line
14
12
10
8 y = 8.3547x0.0443
6
4
2
0
0
25
50
75
100
125
150
175
200
225
Time from Casting (days)
Average for both decks
Power (Average for both decks)
Figure 6.12b. Plots of average CTE for both decks vs time and a trend line
CTE (micro strain/degree C)
6-17
6.6 Deck Induced Girder Deflections Casting a deck atop each girder caused initial deflections due to the dead load (DL), but
the total girder deflections including shrinkage, creep and temperature effects were much greater than the initial deflection. Figures 6.13 and 6.14 show the initial reduction in each girder deflection due to temperature increases in the deck. Following this, as the deck cooled down and began to shrink, the deflection increased to the initial amount and continued much farther. The total deflection for the Type IV was 92% greater at 77 days after deck casting than the initial DL deflection. The BT-56 total deflection at 37 days after casting was 38% greater than the initial amount.
0.80
140
Midspan Deflection (in) Temperature (degrees F)
0.70
120
0.60 100
0.50 80
0.40
60 0.30
40 0.20
0.10
20
0.00 0.01
0.10
1.00
10.00
Time After Deck Casting (days)
IV Midspan Deflection
Average Deck Temp
0 100.00
Figure 6.13 Type IV midspan deflection and average deck temperature versus time after deck casting
6-18
0.80
140
Midspan Deflection (in) Temperature (degrees F)
0.70
120
0.60 100
0.50 80
0.40
60 0.30
40 0.20
0.10
20
0.00 0.01
0.10
1.00
10.00
Time After Deck Casting (days)
BT Midspan Deflection
Avgerage Deck Temp
0 100.00
Figure 6.14 BT-56 midspan deflection and average deck temperature versus time after deck casting
As described in the Instrumentation section, strains were measured in each girder and
deck using vibrating wire strain gauges (VWSG) both before and after casting of the deck. It is
important to note the difference between load related strain and actual strain. For the purpose of
determining prestress losses, a load related concrete strain was calculated using the following
equation.
c = (Ri - R0 ) + T (C1 - C2)
(6.1)
where;
Ri = Strain reading at time i. R0 = Strain reading at "zero" time. T = Temperature difference from reading 0 to reading i. C1 = CTE of VWSG, 12.2 /C. C2 = Appropriate CTE of concrete in /C.
6-19
For these calculations, the zero time was taken as right before cutdown of the
prestressing strands. Using this load related strain in the concrete, a beginning strain profile was
created for each girder before the deck was poured. From this, actual girder strain was added to
obtain a strain profile at some later time. It is important to use the actual strain undergone by
the deck and girder during casting of the deck, as there are considerable temperature changes
taking place. The actual strain was calculated with a similar equation to that of Eq. 6.1, but only
the temperature strain of the VWSG was added as shown below.
c = (Ri - R0 ) + T (C1)
(6.2)
Figures 6.15 and 6.16 contain strain profiles for each composite girder just before casting of each deck and up to 25.5 days and 30 days after casting for the BT-56 and Type IV, respectively. Each figure also contains a cross section of the girders at midspan, to show the corresponding VWSG locations.
70
M2
2.625"
60
7.50"
VWSG Height (in)
50
M4
40
M5
30
20
10
0
-1700 -1450 -1200 -950 -700 -450 -200 50 IV VWSG Actual Strain, Temperature Corrected (micro strain)
Before Deck 7 Days
8.5 Hours 51 Days
29 Hours 92 Days
SOUTH 51"
M6
10.00"
M7 13"
3.00"
TYPE IV SECTION AT MIDSPAN
Figure 6.15 Strain profiles and cross-section of the Type IV at midspan
6-20
VWSG Height (in)
M2
70
3.0"
60
7.25"
50
40
30
20
10
0 -1700 -1450 -1200 -950 -700 -450 -200 50
BT-56 VWSG Actual Strain, Temperature Corrected (micro strain)
Before Deck 7 Days
9 Hours 51 Days
29 Hours
M4 M5
SOUTH
52.75"
M6 M7 11.0" 2.75"
BT-56 SECTION AT MIDSPAN
Figure 6.16 Strain profiles and cross-section of the BT-56 at midspan
Figures 6.17 and 6.18 below show the differences between the average temperature of the deck, girder top flange and bottom flange during and following the casting of each deck. The maximum average deck temperature was 122 oF and maximum M5 thermistor temperature was 105 oF for the Type IV girder. The corresponding maximum temperatures for the BT-56 girder were 117 oF and 102 oF.
The downward deflection of the Type IV and the BT-56 girders in the laboratory closely match the behavior evidenced in the demonstration bridge as discussed in Chapter 5. By 9 hours after the deck was cast when the deck temperature was about maximum, the girder appears to become fully composite. The thermal contraction of the deck then drove the deflection lower than the original DL deflection. Shrinkage of the deck caused further deflection. The strain gauges in the deck show that the deck shrinkage was nearly uniform through the depth. This deck contraction induced girder deflection is a significant factor in overall girder deflection.
6-21
130
120
110
Temperature (degree F)
100
90
80
70
60 0.10
1.00
10.00
Time After Deck Casting (days)
M5 Thermistor
M7 Thermistor
Avg. Deck Thermistors
100.00
Figure 6.17 Plot of temperature versus time for Type IV average deck, IV M5 and IV M7 thermistors
130
120
110
Temperature (degree F)
100
90
80
70
60 0.01
0.10
1.00
10.00
Time After Deck Casting (days)
M5 Thermistor
M7 Thermistor
Avg. Deck Thermistors
100.00
Figure 6.18 Plot of temperature versus time for BT-56 average deck, BT M5 and BT M7 thermistors
6-22
6.7 Transfer Length
Transfer length was measured at each end of each girder, and the ends were designated as IV- W, IV-E, BT-W and BT-E for the Type IV and BT-56 west and east ends. The measurement techniques were identical to those discussed in Chapters 4 and 5.
Figure 6.19 shows DEMEC readings being taken on the Type IV girder.
Figure 6.19 DEMEC reading being taken
The DEMEC measurements were smoothed using a floating 3-point average of strain
values for each girder end, as developed by Russell (1992). The floating 3-point averages were
calculated using the following equation:
x-avg. = x-2 + x + x+2 3
(6.3)
where x-avg. is the 3-point average gauge strain, x is the raw strain at a given point and x-2 and x+2 are the raw strains at 2 in. before and after the location x, respectively. An example of the 3-point averaged, "smoothed" strains is shown in Figure 6.20, for the IV-W.
Figure 6.21 illustrates the 95% AMS Method for the IV-W for the 7-day measurements. The transfer length (lt) was defined by the point where the initial best-fit line intersected the 95% AMS line; In Figure 6.21 the lt equaled 17.4-in. (442 mm).
6-23
3-Point Averaged Strain (micro strain)
1600 1400 1200 1000
Release 3 Day 7 Day 13 Day
800
600
400
200
0
0
10
20
30
40
50
60
Distance from Girder End (in)
Figure 6.20 Smoothed CSS strain data for the Type IV west end
1600 Transfer Length = 17.4"
1400
Average Constant Strain Data
1200 95% AMS Line
1000
Strain (micro strain)
800
Initial Best Fit
Last Data Point for Initial
Line
Best-Fit Line
600
400
200
0
0
10
Smoothed Data
20
30
40
Distance from Girder End (in.)
95% Line
Averaged Plateau Data
50
60
Linear (Smoothed Data)
Figure 6.21 95% Average maximum strain (AMS) method for determining transfer length for Type IV girder, west end at 7 days
6-24
The transfer lengths for each end of the Type IV and BT-56 girders at various days after release are shown in Table 6.4. After 7 days, the transfer length remained constant. The average lt values were taken as the mean of the values from day 7 through day 30.
Table 6.4 Transfer length results for the Type IV and BT-56 girders from the CSS method
Days After Release 1 7 11 16 23 30 Averages
BT West (in) 28 28 27.3 28 27.8 27.7 27.8
BT East (in) 21.4 22 21.4 21 21.4 21.3 21.4
IV West (in) 16.3 18.2 17.4 17.7 18 17.9 17.8
IV East (in) 21 20.5 21.3 20.8 20.7 21 20.9
Day 7 Through Day 30 Averages BT Average = 24.6 IV Average = 19.4
Table 6.4 shows that the average transfer lengths differ from end to end within each girder. The Type IV had an average difference of 3.1 in. (76 mm) and the BT-56 had an average difference of 6.4 in. (163 mm). Past researchers have observed larger transfer lengths at a girder end where there is more "free" prestressing strand extending beyond the girder than at the opposite end of the girder (Russell 1992). This would explain the variation within the BT transfer lengths, as there was approximately seven times the length of free prestressing strand at the west end of the BT versus the east end. This phenomenon does not explain the difference between the transfer lengths of the Type IV girder, as the distance of free strand at each end was essentially the same.
Another factor which is believed to have affected the transfer lengths is the use of thin Teflon mats placed on the prestressing bed at each end of each girder to reduce frictional restraint to girder contraction during transfer. These mats were approximately 1/16-in. thick, 12-in. long and as wide as each girder. They were placed on the steel prestressing bed at the very end of each girder before casting. The top surface of the mat was Teflon while the remainder was a magnetic Vinyl which strongly adhered to the steel bed.
6-25
Slapkus and Kahn (2002) measured transfer length of the demonstration bridge Type IV girders, and obtained an average lt of 27.3 in., but observed cracks in the transfer region of each girder end as shown in Figure 6.22. Slapkus et al. concluded frictional restraint at the girder ends caused the cracking during transfer, and in turn resulted in increased transfer lengths. In the current study, average transfer lengths of 19.4 in. and 24.6 in. were measured for the Type IV and BT-56, respectively. No cracking occurred at the bottom corners of the Type IV or BT56 girder using the Teflon pads. This would indicate the Teflon mats reduced the friction between each girder end and the prestressing bed enough to prevent friction-restraint cracking.
web cracks begin at bottom harped strand and continue into bottom flange
bottom flange cracks begin perpendicular to strands and end perpendicular to bottom of girder
Restraint by the bed
DEMEC strips
Figure 6.22 Typical girder end cracking observed in study by Slapkus and Kahn. (2002)
Table 6.5 provides a comparison between measured transfer length values and predicted values. The three equations used to predict transfer lengths are given in Table 4.4 and were the current equation used by AASHTO and ACI 318 (lt = 50D), the equation presented in the ACI Commentary (lt= (fse)(D)/3) and an equation presented in a study conducted by Mitchell et al. (1993), (Eq. 6.4), which is
3 lt = 0.33 f si D f ci '
(6.4)
The variables used in these equations and their definitions are listed below.
lt: Transfer length
6-26
D: Nominal diameter of prestressing strand fsi: Stress in the prestressing strand immediately after transfer fse: Stress in the prestressing strand after all losses fci': Concrete compressive strength at release
The value of fse was not determined at 28 days, but rather at seven days after casting. This was done to compare with transfer length values determined in a previous research conducted by Slapkus et al. (2002).
Table 6.5 Comparison of measured transfer lengths with predicted values
Source/Equation Used
BT West BT East IV West IV East
Measured Transfer Length (in.)
27.8
21.4
17.8
20.9
Predicted AASHTO, Table 4.4
30.0
30.0
30.0
30.0
Predicted ACI, Table 4.4 Mitchell et al. Insulated fci',
Table 4.4
35
35
36.4
36.4
19.8
19.8
19.0
19.0
Mitchell et al. ASTM fci', Table 4.4 20.7
20.7
19.6
19.6
Insulated indicates accelerated cured for 1 day, then stored in the fog room until testing.
ASTM indicates ambient cured for 1 day, then stored in the fog room until testing.
1 in. = 25.4 mm
Both the equations currently used by ACI 318 and AASHTO, and the equation presented in the ACI 318 Commentary conservatively predicted the transfer lengths. The transfer lengths predicted by AASHTO were on average 18% and 36% greater than actual values for the BT-56 and Type IV, respectively. The equation proposed by Mitchell et al. (1993) provided unconservative transfer lengths for both the Type IV and BT-56 girders.
Transfer length values determined by Slapkus et al. (2002), for Type IV bridge girders averaged 27.4-in., although he observed cracking in the transfer region in each girder end immediately after transfer. Slapkus et al. (2002) assumed the actual transfer length began after the crack location. Using this assumption, the "corrected" transfer lengths averaged 18.4-in.
The transfer lengths for the Type IV measured in the current study were within 5% of Slapkus's corrected, non-cracking average value. The measured average value for the BT-56
6-27
was 33% greater than that found by Slapkus, which can be attributed to the large length of free prestressing strand beyond the girder's west end.
For the Type II girders tested by Reutlinger et al. (2000) and discussed in Chapter 4, the average transfer length was 17.6-in.. This was a 10% and 40% less than the Type IV and BT-56 average transfer lengths measured in this study, respectively. The Type II girders used 8 0.6-in. prestressing strands while the Type IV used 52 such strands. Even though the prestressing force was much greater in the Type IV girders, the transfer length was little different. Therefore, the transfer length results from smaller girders well predicted the transfer length in large girders with flanges full of strands.
6.8 Flexure Test Results and Behavior Each girder was tested twice. The purpose of the first test was to load each girder
beyond cracking, and then unload to study the girder's elastic response. Following this initial test, each girder was then loaded to failure, or near failure to study its inelastic behavior. The test setup is shown in Figure 6.23.
11-in. TYP.
2-W 30x124 SPREADER BEAM
10ft.
1-in. PLATE AND 0.5-in. FIRM
RUBBER PAD
3ft. 4-in. 1ft. 8-in.
1-in. x 12-in. x 26-in. WIDE STEEL PLATE, TYP.
CONCRETE SUPPORT, TYP.
4" STEEL ROLLER, TYP.
87-ft. 4-in. CENTER TO CENTER OF BEARING
Figure 6.23 Drawing of flexure test setup (not to scale)
6-28
The Type IV girder was loaded to failure, which occurred suddenly at 640 kips. The failure was a brittle compression failure of the composite deck, shown in Figures 6.24 and 6.25.
Figure 6.24 View of the Type IV girder flexure failure
Figure 6.25 Close-up of the Type IV girder flexure failure in the deck 6-29
Table 6.5 shows the predicted and experimental moments for the Type IV girder tests. A negative percent difference indicates an unconservative prediction value. The predicted moments at cracking and ultimate were calculated using the AASHTO equations given below;
where; fr: fpe:
Sc: Sb: Md/nc:
Mcr = ( fr + f pe )Sc - M d / nc (Sc / Sb - 1)
(6.5)
Modulus of rupture using 7.5fc' Compressive stress in concrete at girder bottom, due to effective prestress forces only Composite section modulus at girder bottom Noncomposite section modulus at girder bottom Noncomposite dead load moment at section
M n
=
As*
fsu*d1- 0.6
*
f
* su
fc '
(6.6)
where:
Mn:
Nominal moment capacity
:
Capacity reduction factor, equals 1 for flexure
As*:
Area of prestressing steel
fsu*:
Average stress in prestressing at ultimate, (Eq.6.7 below)
d:
Distance from top of deck to centroid of prestressing force
:
Ratio of prestressing steel, As*/bdeckd
fc':
Concrete compressive strength of deck at time of testing
where: fs':
f su*
=
f
s
'
1
-
* 1
*
f
fs c'
'
Ultimate stress of prestressing steel, 270 ksi 6-30
(6.7)
*:
0.28 for low-relaxation steel
1:
1.05-.05*(fc')
(fc' in ksi for deck)
fc':
Concrete compressive strength of deck at time of testing
Table 6.5 Summary of predicted and experimental moments for the Type IV girder
AASHTO Theoretical
Strain Compatibility Theoretical Experimental
Percent Difference
Type IV Moments Mcr (k*in) My (k*in)
95,228
99,337 4.1%
143,070 135,993
-5.2%
Mu (k*in) 148,758
163,601 9.1%
The maximum deflection before failure was 17.5-inches. Figures 6.26 and 6.27 show the load versus deflection plots for both manual and potentiometer deflection measurements.
700 First visible cracks at 363 kips
600
500
Inserted bracing columns 400
Load (kips)
300
200
100
0
0
2
4
6
8
10
12
14
16
18
Manually Measured Deflections (in.)
Cracking Test
Ultimate Test
Figure 6.26 Load versus manual deflection for the initial cracking and ultimate flexure tests for the Type IV girder
6-31
700
600 First Visible cracks at 363 kips
500
400
Ultimate load = 592 kips
Load (kips)
300
200
100
0
0
2
4
6
8
10
12
14
16
18
String Potentiometer Deflection (in.)
Cracking Test
Ulitimate Test
Figure 6.27 Load versus string potentiometer deflection for the initial cracking and ultimate flexure tests for the Type IV girder
The two types of deflection measurements were in excellent agreement which validates the accuracy of the electronic string potentiometer measurements.
During the final flexure test for the BT-56 girder, the strain readings at the level of the bottom strands indicated the girder was rotating laterally. Due to safety concerns, the test was stopped at a maximum load of 456 kips. A check of the girder straightness was performed after the test, and a 0.5-inch horizontal sweep was measured. This sweep would cause the rotation of the girder during loading. Table 6.6 provides the predicted and experimental cracking and yield moments for the BT-56. Figures 6.28 and 6.29 show load deflection plots for both manual and instrument deflections.
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Table 6.6 Summary of predicted and experimental moments for the BT-56 girder
BT-56 Moments
Mcr (k*in)
AASHTO Theoretical
(assumed values)
60,441
AASHTO Theoretical
(actual values)
80,147
Strain Compatibility Theoretical
Experimental
84,336
Percent Difference
(Exp.-AASHTO Predicted)
28.3%
Percent Difference
(Exp.-Strain Comp. Predicted)
My (k*in)
118,543 112,224 -6.4%
Mn (k*in) 130,146
132,354
Load (kips)
500
450
400 First visible cracks at 300 kips
350
300
250
200
150
100
50
0
0
2
4
6
8
10
12
Manually Measured Deflection (in.)
Cracking Test
Final Test
Figure 6.28 Load versus manual deflection for the initial cracking and flexure tests for the BT56 girder
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Load (kips)
500 Inserted bracing columns
450
400 First visible cracks at 300 kips
350
300
250
200
150
100
50
0
0
2
4
6
8
10
12
String Potentiometer Deflection (in.)
Cracking Test
Final Test
Figure 6.29 Load versus string potentiometer deflection for the initial cracking and flexure tests for the BT-56 girder
The initial cracking tests for the Type IV and BT-56 girders showed an elastic response which closely matched theoretical predictions. Each girder recovered most of its deflection; all flexural cracks closed.
The load at which the strain in the bottom most layer of strands reached the strand yield strain (0.010 in./in.) was 532 kips for the Type IV girder and was 443 kips for the BT-56 girder. After yielding, each girder's load-deflection response was plastic; there was little increase in load for significant increases in deflection. This plastic response demonstrated excellent ductility for these girders with bottom flanges fully reinforced with 0.6-in. strands.
During construction, thermal restraint cracks were noted near midspan of the Type IV girder before cutdown, release of the prestressing strands. During testing, flexure and shear cracking did not occur at the locations of the thermal restraint cracks. Those construction, thermal restraint cracks healed as was found by Zia and Hillman (1995). It was concluded that the thermal restraint cracks did not degrade the cracking or ultimate moment strength of the girder.
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At collapse of the Type IV girder, the average strain in the surface of the deck was 0.0029 in./in. This was slightly less than the assumed value given in the ACI and AASHTO standards of 0.003 in./in. This difference is not considered significant because of the excellent ductility of the Type IV girder.
Overall, it was concluded that the AASHTO bridge design standards conservatively predicted the ultimate strength of the Type IV and BT-56 HPC girders with strengths of 14,350 psi and 15,140 psi and with HPC composite decks with strengths of 7,880 psi and 6700 psi, respectively. Further, the standards accurately predicted the cracking strength of HPC girders.
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Chapter 7. Conclusions and Recommendations
7.1 Conclusions
This research project showed that high strength/high performance concrete (HPC) may be used effectively and economically to construct precast prestressed bridge girders. Specifically, it was concluded that HPC bridge girders can be fabricated in Georgia with design compressive strengths (fc') as high as 14,000 psi (96.5 MPa), that highway bridges can be economically constructed using HPC girders and decks, that the demonstration bridge performed well over the more than 3-year monitoring period, that current bridge design standards (AASHTO Standard 17th Edition and AASHTO LRFD 3rd Edition) can be used to conservatively design HPC bridge girders, and that the long-term creep and shrinkage deformations and the long-term prestress losses are less than those predicted using current design standards.
The analytical investigation showed that by using 0.6-in. diameter prestressing strand with concrete strengths up to 14,000 psi, bridge span lengths could be increased up to 40 percent using the same size girder and same girder spacing as compared to the girders made with 0.5-in. strand and 6,000 psi concrete. Concrete strengths greater than 13,000 psi do not help to increase maximum span lengths for AASHTO Type I, II, III or IV girders. Long-span girder stability is satisfactory so long as diaphragms are used to support the girders during deck construction.
The materials studies in the laboratory and at precast plants showed that 7,000 psi, 10,000 psi and 14,000 psi concretes with low permeabilities (less than 2000 coulombs) could be mixed using locally available materials. The laboratory investigations indicated that HPC mix designs should incorporate supplementary cementitious materials (SCM) like fly ash and silica fume in order to achieve HPC's with low to very low permeabilities together with high strengths. The SCM's were needed to develop the required durability characteristics in the Grade 1 through Grade 4 mixes. Field studies showed that HPC's of all grades could be consistently made at precast concrete plants and by ready-mix producers if each producer carefully controlled the mix quality, especially the water content. Both limestone and granite-
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gneiss coarse aggregates from across Georgia were used to make all grades of HPC. The variability of HPC mixes at precast plants was low once quality assurance procedures were adopted.
Creep and shrinkage studies lasting as long as three years showed that HPC had about one-third to one-half the creep of normal strength concrete and that HPC had about one-half to two-thirds the shrinkage of normal strength concrete. The AASHTO specifications provide a good estimate of HPC creep.
The transfer length (lt) of the 0.6-in. strand in 10,000 psi and 14,000 psi HPC was shown to be significantly less than that predicted using ACI and AASHTO standard specifications. The standards predicted transfer lengths between 30 and 38 in. In the Type II girders, the 10,000 psi concrete showed an lt of 17.6 in. (447 mm) while the 14,000 psi concrete showed an lt of 14.6 in.. The 10,000 psi laboratory tested Type IV and BT-56 girders had lt values of 19.4 in. and 24.6 in., respectively.
The development length (ld) of the 0.6-in strands found from testing the Type II girders was 80 in. which was less than the 96-in. length predicted using ACI and AASHTO standards. Therefore, the currently used bridge design standards may be used to conservatively predict strand development length.
The flexural tests of a Type IV girder with 52 0.6-in. strands and of a BT-56 girder with 44 0.6-in. strands showed that the actual ultimate strength of the Type IV girder was 9 percent greater than that predicted using the equations in the AASHTO standards. The cracking moments for the two girders were accurately predicted using those standards; the average experimental value was 2 percent greater than the AASHTO prediction. Overall, the tests of full-size bridge girders made with 10,150 psi (70 MPa) design strength HPC girders and 7,250 psi (50 MPa) HPC composite deck validated that current bridge design standards can be used to accurately predict the cracking and ultimate strength of HPC composite girders.
High performance concrete was used for the construction of precast prestressed girders and of the composite deck for an HPC demonstration bridge in Henry County, Georgia. Type IV girders made with 10,150 psi (70 MPa) concrete were used for the maximum span of 124 ft 1 in. (center-to-center of bearing). The construction showed that the HPC girders could be fabricated with consistently high quality and that the use of HPC resulted in an economical structure. The HPC bridge was constructed at an approximate cost of $51.38 per square foot of
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finished bridge deck, compared with the average cost of $56.65 for a typical, normal strength concrete, precast prestressed composite bridge in Georgia.
The demonstration bridge was monitored for over a three year period. Three months after the deck was placed on the girders, the change in strains at the level of the bottom prestressing strands was insignificant. Therefore, there was little loss in prestressing force due to creep and shrinkage deformations after those three months. The latter was verified through creep and shrinkage tests.
In construction of the 125-ft 3-in. long Type IV girders for the demonstration bridge, cracks were noted near the ends and at the bottom of some girders due to the friction restraint to sliding after cutdown of the strands. Teflon coated, 1/16-in. thick pads were placed at the ends of the Type IV and BT-56 girders used in the laboratory tests. No end friction cracks occurred in these two girders. It was concluded that the Teflon coated pads greatly reduced the friction restraint for these heavy, long-span girders.
Full-depth, thermal restraint cracks were noted in the Type IV girders prior to cutdown of the prestressing strands for both the demonstration bridge and laboratory test girders. The flexural test of the Type IV girder showed that flexure and shear cracking did not occur at the location of these thermal restraint cracks. These cracks which occurred during construction healed and did not degrade the cracking or ultimate moment strength of the girder.
In conclusion, the research project clearly demonstrated that high performance concrete using Georgia materials can be used to fabricate high quality precast prestressed girders for the construction of economical, long-span bridge structures.
7.2 Recommendations for Design Implementation
High performance concretes with design strengths up to 13,000 psi (90 MPa) should be considered for use by the Georgia Department of Transportation. Current AASHTO design standards, both the Standard 17th Edition and the LRFD 3rd Edition, may be used for safe and conservative design of HPC bridge girders. It is recommended that the yield stress for reinforcing bar stirrups not be taken greater than 60,000 psi (414 MPa) as specified in the bridge design standards.
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For HPC girders, the cracking moment should continue to be based upon a tensile stress at the extreme tension fiber of 7.5fc'. While the modulus of rupture (fr) in HPC ranged between 10fc' and 12fc', the actual cracking of full size girders occurred at the lower, recommended value.
The material specifications for HPC should continue to use compressive strength, permeability, curing temperature, and workability standards. Match-cured 4x8-in. cylinders should be used to determine early-age compressive strength. For construction of long-span girders, it is recommended that Teflon coated magnetic pads be used at each end of the girder to minimize friction between the girder and the bottom form in order to eliminate tensile, friction cracking in the bearing-transfer length region of the girders.
7.3 Recommendations for Future Research
Thermal contraction and shrinkage of the composite decks in the bridge and in the laboratory test girders caused increased deflection of the girders. While it would be recommended that such deck-induced deflections be considered when planning the final profile of the bridge surface, there is a lack of data on the magnitude of such deflections for a range of bridge spans and of deck concrete mixes and curing conditions. Therefore, further study of deck-induced deflections is recommended.
Curing temperatures within the bottom flanges of HPC girders reached temperatures over 180oF due to the high cementitious content of the HPC mixes. The use of pozzolonic materials or other supplementary cementitious materials in suitable amounts in HPC may prevent concrete deterioration mechanisms like delayed enttringite formation (DEF). However, precast concrete producers are now fabricating high strength girders using cement-only mixes. It is recommended that DEF be investigated in high strength/high performance concretes to determine if material specifications should require use of supplementary cementitious materials for HPC mixes and/or if those specifications should alter their maximum curing temperature conditions. Additionally, high cement content HPC which do not contain supplementary cementitious materials may also be vulnerable to damage by alkali-silica reaction. Further research is also recommended to assess the implications of the use of cement-only mixtures on durability and the effects of the use of supplementary cementitious materials.
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Construction of low permeability, low shrinkage concrete for the demonstration bridge deck was difficult. Further research into economical mix designs for low permeability, HPC bridge decks is recommended; that research should focus on easy-to-use techniques applicable for ready-mix producers outside the metro Atlanta area. Further, that research should incorporate education of those ready-mix producers.
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References
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ASTM C-39, (1995), Standard Test Method for Compressive Strength of Cylindrical Concrete Specimens, Annual Book of ASTM Standards, Volume 04.02, American Society for Testing and Materials, Philadelphia, PA, pp. 17-21.
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ASTM C-494, (1995), Standard Specification for Chemical Admixtures for Concrete, Annual Book of ASTM Standards, Volume 04.02, American Society for Testing and Materials, Philadelphia, PA, pp. 254-262.
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ASTM C-512, (1987), "Standard Test Method for Creep of Concrete in Compression, "Annual Book of ASTM Standards, Vol. 04.02, American Society for Testing and Materials, 1987
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ASTM C-666, (1995), Standard Test Method for Resistance of Concrete to Rapid Freezing and Thawing, Annual Book of ASTM Standards, Volume 04.02, American Society for Testing and Materials, Philadelphia, PA, pp. 320-325.
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